Guidelines for Design of Dams for Earthquake

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guidelines for design of dams for earthquake

AUSTRALIAN NATIONAL COMMITTEE ON LARGE DAMS

GUIDELINES FOR DESIGN OF DAMS FOR EARTHQUAKE

AUGUST 1998




AUGUST 1998

IMPORTANT DISCLAIMER "ANCOLD and its Members, and the Convenor, Members and Assistants of the

Working Group which developed these Guidelines do not accept responsibility for the consequences of any action taken or omitted to be taken by any person, whether a purchaser of this publication or not, as a consequence of anything

contained in or omitted from this publication. No persons should act on the basis of anything contained in this publication without taking appropriate professional advice in relation to the particular circumstances".

ANCOLD Guidelines for Design of Dams for Earthquake

TABLE OF CONTENTS Page No.

FOREWORD ANCOLD WORKING GROUP MEMBERSHIP LIST INTRODUCTION 1 EARTHQUAKES AND THEIR CHARACTERISTICS 2 2.1 Earthquake Mechanisms and Terminology 2 2.2 Earthquake Ground Motion 2 2.3 Surface Rupture 3

2.4 Magnitude and Intensity 3 2.5 Changes to Seismic Waves Near the Ground Surface 4

2.6 Attenuation and Amplification of Ground Motion 4 2.7 Reservoir Induced Seismicity 5

EARTHQUAKE HAZARD IN AUSTRALIA 6 3.1 General 6 3.2 Mechanism of Earthquakes 8

3.3 Earthquake Depths 9 3.4 Evaluation of Seismic Hazard 10 3.5

Attenuation

11

3.6 Maximum Credible Earthquake Magnitude 11 3.7 Estimates of Ground Motion and Response Spectra at a Site 12

3.8 Earthquake Hazard Maps 12

SELECTION OF DESIGN EARTHQUAKE 16 4.1 Definitions 16 4.2 Selection of the Design Earthquake 20 4.3 Selection of the Operating Basis Earthquake (OBE) 32 4.4 Concurrent Load Combinations 32 4.5 Earthquakes Induced by the Reservoir 33 4.6 Response Spectra and Accelerograms * 33

DESIGN OF EMBANKMENT DAMS AND ANALYSIS OF / LIQUEFACTION 33 5.1 Effect of Earthquake on Embankment Dams 3 3 5.2 General ("Defensive") Design Principles for Embankment Dams 34 5.3 Liquefaction of Dam Embankments and Foundations 36

ANCOLD Guidelines for Design ofDams for Earthquake

6. SEISMIC STABILITY ANALYSIS OF EMBANKMENTS 57 6.1

Preamble

57

6.2 Pseudo-Static Analysis 57

6.3 Simplified Methods of Deformation Analysis 59 6.4 Post Liquefaction Stability and Deformation Analysis 63 6.5 Numerical Methods 65 6.6 Proposed Guidelines 67

7. ANALYSIS AND DESIGN OF CONCRETE DAMS 69 7.1 Past Performance of Concrete Dams in Earthquakes 69 7.2 Defensive Design Measures 70

7.3 Analysis Methods 71 7.4 Design Earthquake and Hydrodynamic Loads 82 7.5 Design Criteria 83 7.6 Dynamic Material Properties 86

8. APPURTENANT STRUCTURES 87 8.1

Introduction

87

8.2 Performance Requirements 87

8.3 Intake Towers 89

REFERENCES APPENDIX A — TERMS OF REFERENCE APPENDIX B — TYPICAL EASTERN AUSTRALIAN PEAK GROUND ACCELERATION VS AEP — RESPONSE SPECTRUM FOR 1 in 1000 AEP — MODIFIED MERCALLI SCALE APPENDIX C — EXTRACTS FROM CANADIAN DAM SAFETY GUIDELINES APPENDIX D — ADDITIONAL INFORMATION ON ACCEPTABLE RISKS


FOREWORD Even in the matter of earthquakes, Australia can be considered the "Lucky Country" in not being on the edge of major tectonic plates. Neighbouring countries like New Zealand and Indonesia are renowned for their volcanoes and frequent earthquakes. Australia is relatively earthquake free by comparison and earthquakes were seldom considered in early dam designs.

Certainly there were some zones of known activity such as the Adelaide Hills and the Western Australian wheat belt, and whilst major damage had occurred, it was not on the same scale as in other countries.

In 1979, the Standards Association of Australia produced the "Earthquake Code" AS2121. It showed zones of seismic activity and recommended methods of determining loads on building structures. The development of this code was based largely on statistics of historic earthquakes, for which there were relatively short term records. However, several major earthquakes subsequently occurred in areas indicated by the code as having

negligible earthquake risk, the most notable being the 1989 earthquake at Newcastle (Magnitude 5.6) in which 12 people died and the Tennant Creek Earthquake in 1988 (Magnitude 6.8). This led to the introduction of a new earthquake code (AS1170.4-1993)which included data from more widespread and reliable seismographs and furthermore considered the all important geological situations. In parallel with these developments, analytical methods used by dam engineers were improving beyond the simplistic application of a horizontal force equating to seismic acceleration. Improvements were

based on the observed fact that earth dams subjected to earthquakes had slumped vertically rather than fail by slipping of a face as indicated by the simplistic analyses. Methods of analysing slumping were developed, and further supplemented by sophisticated finite element analyses which, by utilising modem computer power, give an ability to undertake rigorous analyses of dams where necessary.

This ANCOLD Guideline brings together improved appraisals of the earthquake loadings that a dam may suffer and then describes appropriate methods for analysis and evaluation. Whilst specific to the Australian considerations, the majority of this guideline could be applied to dam structures throughout

the world. The ICOLD Bulletins No. 46(1983), No. 52(1986), No. 62(1988) and No. 72(1989) are parallel documents in this regard, although not including recent advances.

This guideline is a major contribution to dam engineering and the voluntary work by the ANCOLD subcommittee has unselfishly provided their experience to the dam building community and indeed the wider community. Our appreciation goes to Prof Fell and his team for producing this valuable guideline. This guideline is not a design code, and dam designers must continue to apply their own considerations, judgements and professional skills when designing dams to resist earthquakes. As time goes on there will no doubt be improved data and design tools to help the designer and it is intended that this guideline will be updated as circumstances dictate. ANCOLD welcomes contributions to discussion on this guideline which will assist with future revisions.

/

JOHN PHILLIPS Chairman, ANCOLD


MEMBERSHIP OF THE ANCOLD WORKING GROUP FOR

GUIDELINES FOR THE DESIGN OF DAMS FOR EARTHQUAKE 4 Robin Fell

School of Civil Engineering, University of New South Wales

Gamini Adikari

Snowy Mountains Engineering Corporation, Victoria

John Bosler

Snowy Mountains Engineering Corporation, Cooma, New South Wales

Brian Cooper

Dams and Civil Section, Public Works and Services Department, NSW

Peter Foster

Works Consultancy Services, Power Engineering, New Zealand

Gary Gibson

Seismology Research Centre, RMIT, Melbourne

Sergio Giudici

Hydro-Electric Commission, Hobart

Nasser Khalili

School of Civil Engineering, University of New South Wales

Ian Landon-Jones

Dams Safety Group, Sydney Water, New South Wales

Kevin McCue

Australian Seismological Centre, Canberra

Len McDonald

Dams and Civil Section, Public Works and Services Department, NSW

Brian Shannon

Water Resources, Department of Primary Industries, Queensland

David Stapledon

Geotechnical Consultant, Adelaide, South Australia

John Waters

Geo-Eng Pty Ltd, Perth, Western Australia

Ron Wyburn

Halcrow Water Power, Victoria


1. INTRODUCTION

several might have breached, if the reservoir

level had been higher at the time of the

earthquake. Seed (1979), USCOLD (1992), Public awareness of the potential for damage and loss of life in Australia from earthquakes was highlighted by the Newcastle earthquake in December 1989. This was a Magnitude 5.6 (M5.6) event, and because of its proximity to

ICOLD (1986), NSWDSC (1993) and Hinks

Newcastle, and local ground conditions, caused

the frequencies of recurrence of ground

approximately $1 billion damage.

motions. This typically results in 1 in 1000 AEP peak ground accelerations of «0.15g, 1 in 10,000 AEP s»0.35g, and 1 in 100,000 AEP ~0.5g. These are large loadings and it is likely that assessment of many of the existing dams in Australia for such loads could indicate some deficiencies to either the dam or appurtenant

Dam engineers in Australia have been conscious of earthquakes for many years, but it

was the earthquakes at Tennant Creek in 1988, which were M6.3, M6.4, M6.7, with a total fault

scarp length of 32km which raised the question most acutely as to whether dams in Australia could be subject to large earthquakes, and if so, could they withstand them without resultant loss of the facilities and lives, property, and environmental values downstream. Other large

earthquakes in the M6 to M7 range had occurred in Australia, the most notable being in Meckering in 1968 (M6.9), but the Tennant Creek event was critical because it occurred in

an area which had previously been regarded as virtually free of earthquakes. Recent assessments of earthquake ground motions for some large Australian dams have

been based on the assumption that the maximum credible earthquake is M7.5, which is large by any standards. The seismologists involved in these studies indicate that on the available evidence, such earthquakes, ie. M7.5, could occur anywhere in Australia.

and Gosschalk(1993) give some details. The approach taken by seismologists in Australia is to use statistical analysis to predict

structures. New dams would also need (less)

expensive additional design features to cope with earthquake. In recognition of the need to provide some guidance to dam engineers and owners in

Australia, ANCOLD established a Working Group to prepare Guidelines for the Design of Dams for Earthquake. The Working Group was established in September 1993, and took over from an earlier ANCOLD Working Group preparing Guidelines on Seismic Analysis and Design of Embankment Dams. The Terms of Reference for the Working Group are in Appendix A. These guidelines are to cover all types of dams, including tailings dams, and apply to existing and new dams. They cover the selection of the

design earthquake, analysis and design of

In general, it is not possible to identify active

embankment and concrete dams, and appurtenant structures.

faults which might cause such earthquakes. For example, there had been no movement on the

The guidelines are not meant to be used as a

Tennant Creek fault for more than 200,000 years (Crone and Machette, 1992), so the

design code, and of necessity, do not include complete details of all the analysis and design

question arises, can it occur at, or close to any

methods which are recommended. The area is

damsite? Peak ground accelerations close to a M7.5 earthquake can be very high.

rapidly evolving, and those involved in the analysis and design of dams for earthquake should refer to the references given, and to

The past performance of dams in earthquake h&s been very good, with few dams suffering

more recent publications so as to be fully informed. In some situations it will be

major damage. Where this has occurred, it has

necessary to seek specialist advice.

been due to liquefaction in the dam or the foundation. Very few of these dams have breached and released a flood wave. However,


1

2. EARTHQUAKES AND

THEIR CHARACTERISTICS 2.1 Earthquake Mechanisms and Terminology An earthquake is the motion that is produced when stress within the earth builds up over a long period of time until it eventually exceeds the strength of the rock, which then fails and a break along a fault is produced. It may take tens, hundreds or thousands of years for the

stress to build up in a particular area, and it is then released in a few seconds. Part of the energy is transmitted away as seismic waves and part of it as heat.

The fault displacement in a particular earthquake may vary from centimetres up to a few metres in a great earthquake. Once ruptured, the fault is a weakness which is more

likely to fail in future earthquakes, so a large total displacement may build up from many earthquakes over a long period of time. This may eventually measure kilometres for thrust faults produced by compression, or hundreds of

kilometres for horizontal strike-slip faults such as the San Andreas. The point on the fault surface where a displacement commences is called the hypocentre or focus, and the earthquake

epicentre is the point on the ground surface vertically above the hypocentre. The displacement usually propagates along the fault in one direction from the hypocentre, but sometimes it propagates in both directions. Energy release is near but not exactly at the hypocentre.

The hypocentral distance from an earthquake to a point is the three dimensional slant distance from the hypocentre to the point, while the epicentral distance is the horizontal distance from the epicentre to the point.

2.2 Earthquake Ground Motion Earthquake ground vibration is recorded by a seismograph or a seismogram. Most modem

2

seismographs record three components ol motion: east-west, north-south and vertical.

The rupture time for small earthquakes is a fraction of a second, for earthquakes of magnitude 5.0 it is about a second, and for large earthquakes may be up to tens of seconds. However the radiated seismic waves travel al

different velocities, and are reflected and refracted over many travel paths, so the total duration of vibrations at a site persists longei than the rupture time, and shows an exponentia decay.

Several types of seismic wave are radiated from an earthquake. Body waves travel in three

dimensions through the earth, while surface waves travel over the two dimensional surface

like ripples on a pond. There are two types oi body wave (P and S waves), and two types oi surface wave (Rayleigh and Love waves). Primary or P waves are ordinary sound waves

travelling through the earth. They are compressional waves with particle motion

parallel to the direction of propagation. Secondary or S waves are shear waves, with

particle motion at right angles to the direction of propagation. The amplitude of S waves from an earthquake is usually larger than that of the P waves.

P waves travel through rock faster than S waves, so they always arrive at a seismograph before the S wave.

The frequency content of earthquake ground motion covers a wide range of frequencies up to

a few tens of hertz (cycles per second). Most engineering studies consider motion between about 0.2 and 25 Hz.

The amplitude, duration and frequency content of earthquake ground motion at a site depend on many factors, including the magnitude of the earthquake, the distance from the earthquake to the site, and local site conditions. The larger the earthquake magnitude, the greater the amplitude (by definition a factor of ten for each magnitude unit), the longer the


duration of motion, and the greater the proportion of seismic energy at lower frequencies. A small earthquake has low

amplitude (unless it is very close), short duration, and has only high frequencies.

The smaller the distance from an earthquake to the site, the higher the amplitude. The duration is not strongly affected by distance. High frequencies are attenuated by absorption within the ground more quickly than low frequencies, so at greater distances the proportion of seismic

vibration energy at high frequencies will decrease.

2.4 Magnitude and Intensity Earthquakes vary enormously in size. In 1935

Richter defined a magnitude scale to indicate the size of an earthquake. For the Richter local magnitude scale, ML, the logarithm of the peak ground displacement is taken and an empirical correction depending on the distance from

earthquake to seismograph is subtracted. The resulting values are averaged for all the seismographs that have recorded the earthquake.

2.3 Surface Rupture

Other magnitude scales have been defined, including moment magnitude, and while not exactly the same as the Richter local magnitudes, they give similar values that can range from 0.0

Surface rupture is a relatively rare phenomenon which occurs when a fault break reaches the ground surface. It may produce a vertical or

to over 9.0. For each unit of magnitude there is a tenfold increase in ground displacement, and a thirtyfold increase in seismic energy release.

horizontal offset (or both) with a displacement of millimetres to a few metres, and a length

Another measure of earthquake size is the fault

from metres to tens of kilometres.

area, or the area of the fault surface which is ruptured. The fault area ruptured in an

Because rock near the surface is relatively weak, few earthquake hypocentres occur in the top one or two kilometres. It is common for

earthquake depends on the magnitude and stress

surface sedimentary rocks to be folded in response to faulting at depth, giving a monocline and scarp at the surface, but without a surface fault.

drop in the earthquake. For a given magnitude,

a higher stress drop will give a smaller rupture area. Typically, a magnitude 4.0 earthquake ruptures a fault area of about 1 square

kilometre, magnitude 5.0 about 10 square kilometres, and magnitude 6.0 about 100 square kilometres (perhaps 10 by 10 kilometres).

Most earthquakes, especially most larger earthquakes, occur on existing faults. This is because faults are weaker than surrounding

Earthquake Intensity is a measure of the effect

unbroken rock, and are much more likely to fail

normally given on the Modified Mercalli Intensity scale, a copy of which is attached in Appendix B. This is an arbitrary scale defined by the effects observed (whether sleeping

again when stress rebuilds.

A site will have surface rupture potential if an existing fault is found which has been active in the recent geological past (perhaps the past few million years). This will be quite rare, and possibly be difficult to establish. It will usually be easier to show that a site with simple surface geology has no faulting history, than to show that a site with complex geology has suffered recent faulting.

of the seismic waves at the surface, and is

people were woken, trees shaken, etc) and on

the amount of damage caused. Normally the maximum intensity occurs near the epicentre of

the earthquake, and intensity then decreases with distance. However, this may be affected by the orientation of the earthquake rupture, or by local ground conditions such as topography or surface sediments. The earthquake recurrence or seismicity

(seismic activity) of an area must take the range of earthquake sizes into account. There are many more small earthquakes than large. In


3

most places around the earth there are about ten times as many earthquakes exceeding magnitude 3.0 than there are exceeding magnitude 4.0, and ten times as many again exceeding magnitude 2.0. In seismicity studies,

the logarithm of this factor is called the b value, so a value of 1.0 is typical. The b value may be 1.3 or higher if there are many small earthquakes, or 0.7 or lower if there are few small earthquakes.

2.5 Changes to Seismic Waves Near the Ground Surface

2.6 Attenuation and Amplification of Ground Motion Earthquake ground motion attenuates with increasing distance from the source due to

radiation and hysteretic damping. High frequency motion is attenuated more quickly with distance than lower frequency motion. For estimates of peak ground acceleration,

attenuation is allowed for by using an attenuation function of the form a = b,eb2Mir*3

The energy in seismic waves depends upon

where a = acceleration

their amplitude and the physical properties of the material through which they are passing. When waves pass from high stiffness material (eg. rock at depth) into lower stiffness material

vary considerably over the

(eg. near-surface rock, or sediments) they are

world.

reflected towards the vertical and their amplitude increases. Their amplitude also increases as they approach the earth's (free)

R = focal distance M = Magnitude b^jbj are constants, which

bedrock surface can give complex surface

Some earthquake hazard studies use the Esteva and Rosenblueth (1969) attenuation functions, which give peak ground velocity (mm/s), peak ground acceleration (mm/s2) and Modified Mercalli Intensity (IMM) at an epicentral distance x kilometres from an earthquake at depth z kilometres with local magnitude M. The

amplification that varies with earthquake wave

equations are:

surface, at which they are reflected. The nature

and extent of free surface amplification varies with topography, even in fresh, strong rock.

Changes in soil thickness above an irregular

duration. Resonance in the surface sediments causes

amplification at particular frequencies, especially at the natural frequency of the sediments. This depends on the thickness and elastic properties of the sediments. Earthquake motion recorded on hard rock includes all frequencies up to a value that depends on magnitude, while that recorded on soft sediments is usually dominated by the resonant frequency.

In surface sediments, high frequency vibrations are attenuated much more with distance than low frequencies. If sediments are very thick, much of the high frequency motion will be lost and peak surface accelerations will be low, even

if resonance has amplified motion at the low resonant frequency.

4

R Vpeak

Speak

Imm

=

Vx2 +z2 +400

=

160 e10 M R"17

=

20000 e0 8 mR20

=

loge(2980e15MR-23)

Because of the 400 term in the expression for R, corresponding to a minimum R of 20 kilometres, these equations give low values of ground motion at distances closer than a few kilometres.

These relations were determined using Califomian data, and should only be used with magnitudes determined using a compatible function. If the magnitudes computed for seismographs at different distances vary, then the attenuation function is invalid for the area.


In selecting attenuation relationships care is needed, and attention paid to the mechanism of the source earthquake, eg. whether shallow intraplate, or deep crustal boundary earthquakes.

occur further from the reservoir. This occurs at a rate of something like one kilometre per year. RIS is experienced under new reservoirs,

Weak surface materials absorb seismic energy

the stress field and the pore pressure fields

rather than transmit it unchanged, thus tending

under a reservoir have stabilised, then the

to reduce amplitudes at the surface. The amount of attenuation depends on the properties

probability of future earthquakes reverts to a value similar to that which would have existed if the reservoir had not been built. Most of the

of the materials, and especially on their thickness.

Near-surface layers will vibrate preferentially at

usually starting within a few months or years of commencement of filling, and usually not lasting for more than about twenty years. Once

earthquake energy does not come from the reservoir, but from normal tectonic processes. The reservoir simply acts as a trigger.

their own natural frequencies, depending on

their thickness and elastic properties. The

In areas with horizontal tectonic compression

earthquake motion at the natural frequencies of the near-surface layers is amplified, while

and reverse faulting, like Australia, filling a

motion at other frequencies may be little affected or even attenuated. The amplification effect can be especially pronounced for deep soft sediments such as those underlying Mexico City, but in deep, stiff sediments subject to high frequency earthquake, attenuation may result.

Dams (like all other structures) have natural frequencies of their own depending on their mass and stiffness, usually in the range from about 0.5 hertz to about 5 hertz for embankment dams and 2 hertz to 20 hertz for concrete gravity dams.

2.7 Reservoir Induced Seismicity Reservoirs may induce seismicity by two mechanisms. Either the weight of the water may change the stress field under the reservoir, or the increased ground water pore pressure may decrease the stress required to cause an earthquake. In either case, reservoir induced

seismicity (RIS) will only occur if relatively high stresses already exist in the area. If the stress has been relieved by a recent large earthquake, say in the last few hundred years for low seismicity areas like Australia, then RIS is unlikely to occur. /

RIS events initially usually occur at shallow depth under or immediately alongside a reservoir. As years pass after first filling, and groundwater pore pressure increases permeate to greater depths and distances, the events may

reservoir should increase the vertical minimum principal stress and reduce the chance of an earthquake under the reservoir. This has been called reservoir induced a seismicity. However, in some cases earthquakes could then be

induced by later releasing water from the reservoir. Alternatively the change in stress

during filling could induce earthquakes beside the reservoir rather than under it, although this stress change is less pronounced.

It has been suggested that filling a reservoir will cause compression under it, increasing the pore pressure of the existing groundwater, and so tend to induce earthquakes even in areas of horizontal compression. Stress change induced

seismicity, either direct or through this indirect mechanism, should occur soon after filling. It may then cause seasonal variations in

seismicity, sometimes lagging a few weeks or months behind water level.

Pore pressure induced seismicity is normally delayed, and may occur years after filling. Pore pressure increases always tend to induce events.

If there is a major fault near the reservoir, RIS can produce earthquakes exceeding magnitude 6.0 (Xinfengjiang, China, 1962, M6.1; Koyna, India, 1967, M6.3). Such events will only occur if the fault is already under high stress. A number of Australian reservoirs have triggered earthquakes exceeding magnitude 5.0 (Eucumbene, 1959, M5.0; Warragamba, 1973, M5.0; Thomson, 1996, M5.2).


5

MPP*8-

It is more common for a reservoir to trigger a large number of small shallow earthquakes,

especially if the underlying rock consists of jointed crystalline rock like granite (Talbingo, 1973 to 1975; Thomson, 1986 to 1995). These events possibly occur on joints rather than established faults, so are limited in size, and only give magnitudes up to 3 or 4. There is no hazard from such low magnitude reservoir

induced earthquakes, even if they occur regularly. Their shallow depth means that they may often be felt or heard. RIS has been observed for over one hundred reservoirs throughout the world, and small shallow induced events have probably occurred under many others. A relatively high proportion of reservoirs with RIS seismograph networks do record such activity. A high proportion of RIS examples occur in intraplate areas, with above average rates in China,

Australia, Africa and India.

3.1 General The Australian continent is within a tector plate shared with Southern India, so all of earthquakes are intraplate. The pla boundaries to the north and east are among t most active on the earth. Possibly as a result

this, Australia is one of the most actr intraplate areas on the earth. Despite this, tf hazard is quite low when compared with acth interplate areas.

Most people in Australia can expect to feel earthquake about every five or ten year although many of these may not be recognisf as an earthquake. Most Australian earthquake

that are reported are heard with a noise like distant quarry blast or thunder, with possibly slight vibration being felt. Only a proportion of earthquakes that are fel perhaps about one in twenty, will cause som

It is not easy to predict whether a reservoir will experience RIS because the stress and strength at earthquake depths are not easily measured. For the same reason, prediction of normal tectonic earthquakes has been unsuccessful in most parts of the world.

It seems that RIS with many small events is more likely to occur in intraplate areas with near surface crystalline rocks like granite, rather

than sedimentary rocks. A larger magnitude RIS event can only occur if there is an existing fault of sufficient dimension that is late in its earthquake cycle (the stress is already approaching the strength of the fault).

3. EARTHQUAKE HAZARD IN AUSTRALIA The understanding of the hazard imposed by earthquakes in Australia is critical to selection and application of design earthquakes for dams. Hence, a relatively detailed discussion on the topic follows. This is largely taken from Gibson

(1994).

damage in their epicentral area. If they occur i an inhabited area, most earthquakes larger tha about magnitude 4.0 will cause some damage.

By contrast, in an active interplate area lik New Britain or Bougainville in Papua Nev Guinea, earthquakes are felt very often, c average every week or two. These are normal

felt rather than heard, with any sounds being thf reaction of a building to the vibration rathe than the earthquake itself. A very small proportion of these PNC earthquakes, perhaps about 1 in 500, havf caused any damage in their epicentral area, an<

earthquakes smaller than magnitude 6.0 rarelj cause damage.

There are a number of factors that influence the estimation of earthquake hazard in Australia. One is the short duration of documented history, with a little over 200 years of data about Sydney and considerably less for most of the continent.

Another factor is the large area of the country relative to the size of the population. Seismographs are distributed relatively sparsely, limiting the accuracy of earthquakej locations.


As was the case over most of the earth, seismograph coverage of local earthquakes in

Organisation (AGSO), aims to locate all earthquakes in Australia larger than magnitude

Australia was only established in about 1960 following the International Geophysical Year. There should be some link between population and seismograph density with more

ML3.0. Local seismograph networks have non-uniform coverage, often with good coverage of large dams and poor coverage of rqajor cities.

instrumentation in the populated southeast. However, coverage is still far from complete.

Figure 1 shows the location of Australian earthquakes with magnitude exceeding ML4 in the period 1850 to 1993 (note that many earthquakes prior to 1960 will not be shown).

The Australian National Seismograph Network, operated by the Australian Seismological Centre at Australian Geological Survey

Table 1. Some Large Australian Earthquakes (Gibson, 1994).

1

Date 1892-01-26

Place Flinders Island, TAS

2

1897-05-10

Beachport, SA

No.

Mag

Imax

Damage, A$(1990) Where Given

6.9

7+

Offshore epicentre, little damage. Felt throughout Tasmania and Eastern Victoria

6.7

8

Damage in Kingston, Robe and Beachport,

liquefaction At Warooka chimneys fell, walls partially demolished, few buildings without damage

3

1902-09-19

Warooka, SA

6.0

8

4

1903-07-14

Warrnambool, VIC

5.3

7

Extensive minor damage and liquefaction at Warrnambool. Followed similar event April

5

1918-06-06

Bundaberg, QLD

6.0

6

Epicentre offshore, minor damage in

6

1941-04-29

Meeberrie, WA

7.2

8

Isolated area. Damage to remote farm houses included cracked walls, burst tanks

7

1954-02-28

Adelaide, SA

5.4

8

Widespread minor damage. $71m. Epicentral area was rural, but is now urban

8

1961-05-21

Robertson, NSW

5.6

7

Damage in the Moss Vale, Robertson and Bowral area. $3.4m

9

1968-10-14

Meckering, WA

6.8

9

Most buildings in Meckering destroyed.

10

1969-06-20

Boolarra, VIC

5.6

6

Rockhampton

$29m. 32km surface rupture

Cracked walls and fallen chimneys in epicentral area

11

1973-03-09

Burragorang,NSW

5.0

6

Minor damage in Picton, Bowral and Wollongong. $2.3m

12

1979-06-02

Cadoux, WA

6.2

9

Many buildings at Cadoux destroyed, most others damaged. $9m. Only one injury. Surface rupture extended over 15km

13

1986-03-30

Marryat Creek, SA

5.9

8

Epicentre in remote area. Cracked walls in nearest homestead, 13km surface rupture

14

1988-01-22

Tennant Creek, NT

6.8

9

Epicentre in remote area. $1.2m, mainly to

15

1989-05-28

Uluru, NT

5.7

6

/ 16

1989-12-27

Newcastle, NSW

5.6

8

damaged gas pipeline, 35km surface rupture Minor damage at Uluru National Park (Ayers Rock). Epicentre west of Mt Olga Thirteen fatalities, $1500m + damage. Widespread minor to moderate damage

17

1992-09-30

Amhem Land, NT

5.1

5

Epicentre offshore. No significant damage


1

principal stress near vertical, producing revet or thrust faults. There may be some strikes

3.2 Mechanism of Earthquakes

movement, especially if the failure is on an { fault that is oriented at an angle to the princij

Almost all Australian earthquakes have mechanisms with the maximum principal stress

stress.

near horizontal. Most have the minimum

113

123

120

133

130 ro—i-

u 0s ^ O

10 no <&> 0 0

150

143

140

O00 9 &£> o

o-Ofyo0 a a O O

aO q.

O do o

15

*6 If 90 o . Q © 20

a

*

s> 8 o

Q o 00

o

a!? o CO • o o

40

Australian Earthquakes to 1994

H

Magnitudes: *4 05 o

Figure 1. Australian earthquakes with magnitude exceeding ML4.0 since 1850 (Gibson, 1994). Earthquakes on reverse faults usually give a high stress drop, where the seismic energy

Compression giving reverse and thrust faul produces surface uplift. Therefore, earthquakf

comes from a small source volume. High stress

are most likely to occur in areas whei

drop earthquakes radiate a greater than average proportion of their energy in higher frequencies. Higher frequency motion implies higher

mountains are developing, and less likely und< large sedimentary basins. They may b expected under the Eastern Highlands an Flinders Ranges, but are less likely under th Murray Basin, Great Artesian Basin and tl

accelerations for a given energy release or

earthquake magnitude, but not necessarily fewer cycles, than for similar magnitude

Eucla Basin.

earthquakes in, say, West Coast USA.

8


It may be argued that dams are particularly likely to be built near to an active fault. This is because dams are built in valleys which have

Eastern Australian earthquakes are usually at

recently been eroded. Erosion requires uplift,

deeper. Australia's deepest known earthquake

and faulting is the geological process that gives

occurred at 39km depth offshore from Arnhem Land in 1992 (McCue and Michael-Leiba,

most pronounced uplift. In places where horizontal compression produces reverse

depths between 1 and 20km, while those in Central and Western Australia may be a little

1993).

faulting, the fault will inherently dip back under the upthrown block, and under the dam. Unlike at plate boundaries, there are no large dominant active faults in Australia. Instead,

In Eastern Australia, earthquakes at depths of less than 5km are regarded as shallow, and greater than 15km are regarded as deep. The

earthquakes are distributed over many smaller

Newcastle earthquake of December 1989 was at a depth of about 12km, and a number of

faults. Most of these have poor surface outcrop

earthquakes about Katoomba were at about 18m

and have not yet been identified. Large earthquakes, eg. Tennant Creek (1988), have

depth.

occurred on faults which were not previously apparent on the surface. Crone and Machette

All of these Australian earthquakes are very shallow compared with those in plate boundary

(1992) describe evidence that there had been no

areas, where earthquakes can occur as deep as

movement on the fault at Tennant Creek for 200,000 years.

700km, and depths of less than 70km are regarded as shallow, and greater than 300km as deep.

Based on the available information, it would appear that the differences between Australian earthquake ground motion and those from

It has been suggested that small earthquakes are more likely to be shallow, while larger

earthquakes in USA, China etc, on which many

earthquakes can occur at all depths.

of the dam design methods are based, are not

sufficiently great as to invalidate the design methods. This is an aspect which will need to be further assessed as more ground motion data

for Australian earthquakes is gathered.

3.3 Earthquake Depths Usually, earthquake depths can be precisely determined only if the distance to the nearest seismograph is not greater than about twice the earthquake depth, or about 10 kilometres. The depths of most Australian earthquakes are poorly constrained, because of their relatively shallow depths and the low density of seismographs.

If it is not possible to constrain an earthquake depth using seismogram data, an arbitrary "normal" depth may be used. The value of the

normal depth may or may not be realistic, and typical values used in Australia include 0, 5, 10 or 33 kilometres.

Table 1 shows some large Australian

Shallow earthquakes often have many aftershocks, some with magnitudes approaching that of the main shock. Deep earthquakes usually have few aftershocks, and these are often much smaller than the main shock.

Because of their shallow depth, small Australian earthquakes are often felt or heard. Magnitude 1.0 events can be felt to a distance of about one kilometre and magnitude 2.0 to about four kilometres. These are slant distances, so

only very shallow events of these magnitudes will be felt. Because the short travel distances do not give very much attenuation of high frequency vibrations, and because a high stress drop gives a high proportion of high frequency vibration from the earthquake source anyway (see Section 3.7), these events have enough energy

hi the audio frequency range to allow them to be heard. For many such small earthquakes the sound heard is more significant to an observer than the vibrations felt.

earthquakes and a description of the damage.


9

For similar reasons, moderate magnitude earthquakes in Australia produce motion with strong peak ground accelerations, and can produce significant damage. The Newcastle

earthquake was only of magnitude MLS.6, but caused extensive damage.

3.4 Evaluation of Seismic Hazard The estimation of ground motion requires the following seismicity information about the surrounding area: • the rate of occurrence of earthquakes • relative proportion of small to large events

(b value) • maximum earthquake size expected

(maximum credible magnitude) • the spatial distribution of earthquake epicentres including delineation of faults. The seismicity can be evaluated by a modified

Richter relation (Gibson, 1994) log,o(P) = - log10[ 10( 10"bM-10"bMmax] -log10(No)

P is the return period in years for an earthquake of magnitude M or greater. N0 is the rate of occurrence of earthquakes,

given as the number of earthquakes of magnitude zero or greater per year per unit volume or per unit area. An area

of 100 x 100 kilometres is commonly used. This must be converted to per

square kilometre or per cubic kilometre for ground motion calibrations.

b is the Richter b value, which gives the relative number of small earthquakes to large. It is the logarithm to the base 10 of the ratio of the number of events exceeding M+l to the number exceeding M. A value of 1.0 would correspond to ten times as many

earthquakes exceeding magnitude M as would exceed magnitude M+l.

;/

infinite return period. Because of t

low probability of very lar earthquakes, Mmax does not critica affect ground motion recurren estimates for return periods up

hundreds of years, especially when t b value is high. However, it is mo important for low AEP events such may be important for design of hi| hazard dams. The maximum credit magnitude causes the magnitin recurrence plot to flatten out ai asmyptote to that value.

The seismicity parameters N0 and b determined using available earthquake data In most places there are insufficiei earthquake data to determine Mmax fro historical records, but values can be estimate

by considering the tectonic situation and loc, fault dimensions. The b value is much moi critical than N0, and a small adjustment to will give a large change in N0. Hazard evaluation depends on tfi extrapolation of data from smaller earthquake to larger magnitudes. The lower the b valui the greater the resulting hazard estimate, catalogue with missing small earthquakes wi give an invalid low b value, and an estimate hazard that will be too high. A catalogue wit smaller quarry blasts incorrectly identified a earthquakes will give a high b value, and ai extrapolation that is non-conservative.

In Southeastern Australia the b value varie from about 0.75 in Western Victoria to a littl over 1.0 south of Sydney, then down to 0.90 ii Northeast New South Wales. Tto corresponding N0 value is about 0.002 event larger than MLO.O per cubic kilometre per year, These methods are commonly used b; seismologists to estimate design earthquakes ii Australia. Their advice is that they an reasonably reliable, when combined witl suitable attenuation functions, and can be use<

to estimate ground motion for low AEP events-

Mmiix is the magnitude of the maximum

credible earthquake for the area. It is the magnitude of an earthquake with

10


3.5 Attenuation The accelerograph and seismograph data currently available are not sufficient to allow determination of good local attenuation functions in Australia, especially spectral attenuation functions.

If a valid attenuation function is being used, magnitudes computed from seismograph data will not vary with distance. When the Richter attenuation function from California is used in Southeast Australia there is no significant variation with distance. If it is used in Central or Western Australia, magnitudes increase with distance. This means that attenuation in

Southeast Australia is similar to that in California (a little above the world average), while in Central and Western Australia it is well

versus AEP. For this, it is necessary first to assess the magnitude of the maximum credible earthquake. If the earthquake catalogue only covers a short

period compared with the required return period, then a large area must be considered when estimating Mmax, perhaps as large as the

whole of Australia or even including other intraplate areas over the earth.

Table 1 lists some of the largest earthquakes which have occurred in Australia. It will be seen that there are several in the range M6.5-M7.

To give some appreciation of the likely maximum credible magnitude, seismologists have considered the credible length and depths of fault which may rupture to cause an earthquake.

below the world average.

It follows that the Esteva and Rosenblueth equation is likely to be reasonable for estimating acceleration in Eastern Australia, but

will underestimate peak ground accelerations within 20km focal distance of the earthquake. The Gaull et al (1990) equation should be used to estimate velocity.

Gaull et al (1990) have used the variation of Modified Mercalli Intensity with distance to estimate attenuation functions for Western, Southeastern and Northeastern Australia. It is

suggested that these estimated attenuation functions may be used to estimate peak acceleration and velocity in these regions. It is pointed out that care is needed in all these calculations because the earthquake magnitudes have been estimated from assumed attenuation functions, so one needs to be consistent. Where

it is critical, it is clear that the advice of seismologists should be sought on attenuation, as well as other aspects of earthquake loadings.

3.6 Maximum Credible Earthquake Magnitude /

In view of the impracticabilityof identifying the faults in Australia on which major earthquakes may occur, it is necessary to use probabilistic methods to estimate expected ground motion

There is an approximate relationship between the rupture area of a fault and the magnitude of the earthquake (Gibson, 1994). Magnitude

Fault Area km2

4 5 6 7

1 10 100 1000

If earthquakes are constrained to shallow depths, as is the case in Australia, then a large

earthquake will require a very long fault. For example, if the down-dip distance of a magnitude 7 earthquake is 20km, then the strike length will be 50km. A magnitude 8 earthquake similarly constrained would give an unlikely rupture length of hundreds of kilometres. Very few intraplate earthquakes are larger than magnitude 7.5. A series of very large intraplate earthquakes in the New Madrid area of Missouri, USA, in 1811 to 1812, apparently had magnitudes exceeding 8.0 (Johnston and Shedlock, 1992). Other authors believe that the New Madrid earthquakes were smaller (Everhden, 1975, M6.9; Gomberg, 1992, M7.3; Gibson, 1994). There is a general consensus among Australian

seismologists that the maximum credible


11

magnitude of an earthquake in Australia is around M7.5. This is based on consideration of

the depth at which earthquakes occur and the sensible length of fault which will rupture. A

Response spectra are estimated using data fr< overseas earthquakes, modified to §

Australian conditions. Data from Austral! earthquakes are less reliable because of t

M7.5 earthquake could occur anywhere in the

scarcity of data. Typical peak grou

country.

accelerations vs AEP and response spectra t

given in Appendix B, but an individi Given the shallow seismogenic depths in

assessment should be made for each dam.;

Eastern Australia, the value of 7.5 for Mmax is

should be noted that these will usually be f bedrock conditions, unless otherwise specifiec

possibly conservatively high. For return periods to 1000 years, decreasing it to 7.2 or 7.3 would have negligible effect on ground motion recurrence estimates. For long return period

The installation of digital accelerographs If shown that Australia experiences high pe

motion, the lower value could be justified, especially if there is little earthquake and geological evidence for large nearby faults. If the greater depth of Central and Western Australian earthquakes is verified, the value of 7.5 may not be conservative. Gibson (1994) indicates that there is some evidence that earthquake activity is cyclical, occurring in

ground accelerations. An ML4.9 aftershock the Tennant Creek earthquakes was recorded

clusters, perhaps about every hundred years for active areas, or at much longer intervals for

structure. Accelerations with higher th,

inactive areas. Clusters are affected by activity in neighbouring areas, so at any particular place the activity is cyclical rather than periodic. The fact that say a M6.5 earthquake has occurred does not preclude a M7.() or M7.5 in the foreseeable future (ie. <100 year). In fact, there is some evidence that large earthquakes may be

intraplate areas such as Canada.

preceded by smaller ones.

a distance of 9km with a peak grow acceleration of 0.53g, at high frequency. Mc accelerograms recorded in Australia to da show these high accelerations. These hi; accelerations occur in high frequency peai which have little impact on the response of expected values are also being recorded in oth

For other than for very preliminary studie estimates of ground motion and respom

spectra should be made by experienc< seismologists.

3.8 Earthquake Hazard Maps

Given that since 1968 Australia has experienced four earthquakes of magnitude 6.2 to 6.8, all of which produced surface rupture, from 13 km to 35km in length, it does not seem "incredible" that we could experience an M7.5 event.

It is worth noting that when one considers the 10"3 AEP earthquake for say Sydney, one may be expecting a M6 or M6.5 earthquake. By comparison a 10"3 AEP earthquake for San

Francisco will be say M8 (Committee on Safety

Criteria for Dams (1985) suggest M8 has 150 year return period for San Andreas fault). However, when one considers the 10"5 AEP

event in Sydney it is likely to be around M7.5, and San Francisco will have risen to say M8.5. Hence, the differential becomes less at the lower probability events.

3.7 Estimates of Ground Motion and Response Spectra at a Site Estimates of ground motion at a site are made by combining the earthquake occurrence

(a) History of Development Earthquake hazard maps of Australia ha\ evolved considerably over the past eightee years, in line with improved seismograp coverage, more and better earthquake data, an

a better understanding of the geologic! processes producing earthquakes. They may t useful in defining design earthquakes, at lea for high AEP events. Figures 2 to 5 are examples of these map showing the reducing emphasis on particuk past earthquakes and on the seismograp distributed with time. McEwin, Underwood and Denham (1976 produced a series of maps showing the expecte 50 year return period ground acceleration: velocities and intensities. The map giving th 50 year return period peak ground velocities i shown as Figure 2. Data were taken from th period 1960 to 1972. Only six seismograph were operating in Australia before 1960, an

estimates, with a suitable attenuation function.

12

AN COLD Guidelines for Design of Dams for Earthquake

"K

P they were of little value for the study of smaller If local earthquakes. |f The maps produced were significantly affected I by the distribution of seismographs at the time, and by the locations of the large earthquakes that occurred during the period.

Most of Australia was in zone zero, but this did

not imply that earthquakes would not occur in these areas. It was stated that significant shaking may arise in any part of this zone in the future.

Gaull, Michael-Leiba and Rynn (1990) used programs based on the Comell-McGuire

The Standards Association of Australia (1979) produced the SAA Earthquake Code AS2121-1979. This included a map showing four zones of earthquake hazard based on the expected recurrence of peak ground velocity,

Figure 3. It was produced using data up to 1976, and also considering historical earthquake activity from 1897. It was felt that data were complete for all earthquakes exceeding magnitude ML4.0 from 1969. The map was still dominated by the larger earthquakes that had occurred between 1960 and 1976, and was inevitably influenced by the seismograph distribution. However, it did recognise new areas of activity in Queensland and Central Australia. Where uncertainties were very high, zones were

method to produce plots of ground motion with a 10% chance of being exceeded in a fifty year period. These included peak ground velocity, peak ground acceleration (Figure 4), and Modified Mercalli Intensity. The data used covered the period from 1859 to 1987. The maps showed larger areas susceptible to earthquakes, but with higher values for those areas which had experienced a large earthquake within the past 100 years. They show a reducing emphasis on particular past earthquakes, and on the seismograph distribution with time. Between 1986 and 1989, several earthquakes exceeding magnitude ML5.0 occurred in areas that had little or no previous known activity. These included Maryatt Creek in northern

deliberately given simplified rectangular

South Australia (1986), Nhill in Victoria

boundaries.

(1987), Tennant Creek in the Northern Territory

(1988), Uluru in the Northern Territory (1989), Newcastle in New South Wales (1989).

Figure 2. 50 year return period peak ground velocity (McEwin, Underwood and Denham, 1976).


13

130°

.Tv;; • •

Figure3 AS2121-1979- Earthquake zoning.

Figure 4. Peak ground velocity (mm/sec) with a 10% chance of being exceeded in a 50 year perioi (Gaulletal, 1990).

14


Figure 5. Acceleration coefficient map for AS 1170.4-1993.

(b) The 1993 Standards Australia The map is based on the best available

Earthquake Code A Standards Australia Committee began

Australian attenuation functions, derived from isoseismal maps. The numerical values on the

completed over ten years before, and there was

velocity map were consistent with those obtained from other studies. There is better correlation of peak ground velocity with intensity and damage than with peak ground

insufficient time for the earthquake hazard to be computed using either spectral attenuation

acceleration. There is also considerable uncertainty with acceleration attenuation

functions or a new computer based GIS

functions for Australia, and the higher values being measured would need to be adjusted for

working on a new Code for minimum loads on

structures (Standards Australia, 1993) in late 1989. The previous Earthquake Code had been

program.

short duration structural response.

It was decided that hazard was to be represented by a single number varying with location, so it was not possible to distinguish between high frequency vibration hazard (with acceleration dominant) and low frequency vibration hazard (with velocity or displacement dominant).

The map was based on a Comell-McGuire source zone configuration. This consisted of

polygons each with the equivalent of constant N0, b and Mmax values. The polygons used had

sharp boundaries rather than fuzzy boundaries, producing some angular patterns in the

To avoid the problems of major increments at zone boundaries it was decided to produce a

contours. A number of additional earthquakes had occurred.

contour map, and to define standard values for use in major cities.

It was decided to use the velocity map as a base map and to smooth the contours compared to

The hazard estimate is based on the velocity recurrence map of Gaull et al (1990).

the earlier code, according to the collective experience and knowledge of the committee.


15

The smoothing was done after consideration of the gross geology and tectonics of Australia, new information on historical earthquakes, and

the earthquakes that had occurred since the map was originally drawn, eg. particularly the 1988 Tennant Creek and 1989 Newcastle earthquake events.

known as the Operating Basis Earthqua (OBE) and the other for the condition whi severe damage is expected, but uncontrolj release of the storage water is to be prevent!

known as the Maximum Design Earthqua (MDE). The working group has decided, adopt slightly modified versions of { International Commission on Large Dai

The resulting acceleration coefficient map (Figure 5) was smoother than the original, in recognition of the uncertainties involved. A number of contours are plotted to facilitate interpolation, and should not be taken as an indication of precision. The map was converted from peak ground velocity to an "acceleration coefficient" for use

in the code. The acceleration coefficient is a dimensionless quantity numerically similar to peak ground accelerations that would have been computed using traditional attenuation

(ICOLD) definitions (ICOLD, 1989) for OI and MDE, as follows: Maximum Design Earthquake - the ME will produce the maximum level of groui motion for which the dam should designed or analysed. It will be required least that the impounding capacity of t dam be maintained when subjected to th seismic load.

functions. These values are lower than peak accelerations actually being measured from

Operating Basis Earthquake - the 01 will produce a level of ground motif which will cause only minor and acceptab damage at the damsite. The dai

Australian earthquakes, but may be thought of

appurtenant structures and equipme

as being an "effective acceleration" after

should remain functional and damage fro

consideration of the short duration of the strong

the occurrence of earthquake shaking n

motion.

exceeding the OBE should be easi repairable.

The map gives the broad variation in earthquake ground vibration hazard over the country. The effects of site response due to weak surface

materials or topography are allowed for in an additional site factor, in the code. The map does not allow for any effects of local faults, and does not cover hazards other than ground vibration, such as surface rupture or the

potential for landslides or tsunamis. Site response and these other hazards vary over small distances, so could be represented in local earthquake zoning maps, perhaps to a scale of

about 1:100,000.

4. SELECTION OF DESIGN EARTHQUAKE

used in the assessment. The MCE is defined a! Maximum Credible Earthquake - t MCE is the largest reasonably conceival earthquake along a recognised fault within a geographically defined tectoi province, under the known or presum tectonic framework.

ICOLD (1989) allow definition of MCE terms of either magnitude or intensity at t dam site. The Working Group are of the vi( that this can be confusing, and recommend i of MCE in terms of magnitude only. In some situations, where two or more causati

4.1 Definitions Dam engineers, in keeping with engineers who design other structures, have commonly

adopted a deterministic or standards based approach to design, with two levels of design earthquake motion; one for serviceability,

16

When considering the MDE, the concept

Maximum Credible Earthquake (MCE) is oft

faults are identified, each may have an MCE. The MCE which is critical for the design of < dam is known as the Controlling MaxiiW

Credible Earthquake (CMCE).


The MDE is expressed in terms relevant to the

Probabilities can be expressed also in terms of the life of the structure or some other period. In

design of the dam, which may include peak

this case the probability is expressed as the Y

ground acceleration or velocity, earthquake magnitude and distance or response spectra.

year EP, for example, the 50 year Exceedence

Where causative faults can be identified the MDE may often be determined from the MCE (or CMCE). Where such faults cannot be identified, the MDE can be determined by probabilisticmethods.

The AEP and Y-year EP are related by the

It is recommended that the probability that a particular earthquake magnitude or ground motion (eg. peak ground acceleration or

Probability may be 1 in 100.

formula Y-year EP= l-(l-AEP)y

Table 2 lists exceedence probabilities for periods of 50 to 500 years calculated from AEPs varying from 1 in 100,000 to 1 in 100.

velocity) will be exceeded in any year be expressed as the Annual Exceedence

Probability (AEP), and expressed as a 1 in Y terminology, eg. AEP of 1 in 1000.

Table 2. Relationship between Annual Exceedence Probability (AEP) and Exceedence Probability (Y-year EP) for Different Periods.

AEP 1 in 100,000 1 in 10,000 1 in 1,000 1 in 100

Y-year EP for Different Periods Y = 50 Years Y = 100 Years

Y = 200 Years

Y = 500 Years

1 in 200 1 in 20

1 in 2000 1 in 200 1 in 20

1 in 1000 1 in 100 1 in 10

1 in 500 1 in 50 1 in 5

1 in 2.5

1 in 1.6

1 in 1.2

Table 2 shows that the AEP of an earthquake can be a misleading measure of general security, suggesting a greater degree of safety than actually exists over a long period of years. For example, if a dam is safe for an earthquake

which has an AEP of 1 in 100, and if one is interested in the safety of downstream residents over a period of say 50 years, then the 50-year

EP of that earthquake is 1 in 2.5, ie. there is a 40% chance of that earthquake occurring in any 50 year period.

1 in 2.5 1 in 1.01

As will be discussed later, the design of dams for earthquake may be done using risk assessment methods. In this guideline, the

following definitions relating to risk assessment are adopted from the ANCOLD Guidelines on Risk Assessment (ANCOLD, 1994). Hazard:

That which has the potential for creating adverse consequences. Hazard is rated

according to the scale of the adverse consequences.

Thus, in assessments of safety issues, a better

perspective of relative safety is often obtained from consideration of earthquake exceedence

probabilities over a period of years which relates to the life of the structure or the period of occupancy, rather than from consideration of

the AEP.

Hazard Category: A scale of adverse consequences caused by dam failure.

Risk: The likelihood or probability of adverse consequences; the downside of a gamble.


17

Failure Type 1 (Fl): j

Acceptable Risk: That level of risk that is sufficiently low that society is comfortable with it. Society does not generally consider expenditure in further reducing such risks justifiable.

A major failure involving complete abandonment of the dam.

Failure Type 2 (F2): j A failure which at the time may hi been severe, but yet has permitted extent of damage to be successfi

Risk Analysis: A policy analysis tool that uses a knowledge base consisting of scientific and science policy information to aid in

repaired, and the dam brought into again.

For the purposes of this guideline, the te failure will refer to both Type 1 and Typ

decision making.

failures.

Risk Assessment:

The total process making use of risk analysis and which embraces a consideration of all costs and benefits including the identification of risks, the estimation of their likelihood of occurrence and the evaluation of the social acceptability of the risk. Risk assessment includes the processes that lead to

Accident: Three categories of accidents are listed

ICOLD(1983): Accident Type 1 (Al): An accident to a dam which has b in use for some time, but which been prevented from becoming

failure by immediate remei measures, including possibly draw

decisions. NOTE: The definitions above are to be reviewed if/when ANCOLD agrees to any change

down the water.

Accident Type 2 (A2): An accident to a dam which has bi

Risk Management: The reduction or control of risk to an acceptable level.

observed during the initial filling the reservoir and which has bj prevented from becoming a failure! immediate remedial measu

Socially Acceptable Risk: That low level of risk that society finds tolerable so that expenditure would not normally be directed to its reduction.

including possibly drawing down! water.

i

Accident Type 3 (A3): i An accident to a dam du| construction, ie. by settlement

Individual Risk Criterion: The socially acceptable level of risk to a particular individual. Societal Risk Criterion: The socially acceptable level of risk in terms of events that impact on society at a community, regional or national level.

foundations, slumping of side slo etc which have been noted before water was impounded, and where essential remedial measures have b carried out, and the reservoir sa:

filled thereafter. Deterioration:

The definitions of failure, accident, deterioration and incident are taken from

Any faulty behaviour, from the poin view of both safety and performance, ei during construction or after being!

ICOLD(1983):

service, including failure cases.

Failure: Two categories of failure are listed by

ICOLD(1983):

18

Incident: Either a failure or an accident requi major repair.


Total Risk is calculated from

Breach:

A failure resulting in rapid release of water Total Risk = (probability of event) x

from the dam.

(consequencesof that event)

The ANCOLD (1994) definition of risk is not which has the dimension

explicitly linked to the consequences, and is expressed in terms of probability.

consequence event consequence x unit time unit time event

In some situations it is necessary to assess the total annual consequence of failure, and it is

useful to assess the total risk (either loss of life, or economic consequences), where total risk is defined as:

Total Risk: A measure of the probability and severity of

This definition is adapted from that used by the Canadian Dam Safety Association (1994) for "risk". It will be noted that the total risk can be posed by a number of causative events, eg. flood, earthquake, and other, eg. piping failures,

slope instability, in which case the total risk is a

an adverse effect to health, property or the environment.

summation from the causative events.

Total risk is estimated by the mathematical expectation of

For the purposes of this guideline, the hazard

the consequences of an adverse event occurring.

categories are defined as shown in Table 3.

These are taken from ANCOLD (1986) and are under review by ANCOLD. The revised guideline will include assessment of the environmental consequences of dam failure and is likely to have more hazard categories.

Table 3. ANCOLD Dam Hazard Categories (ANCOLD, 1986).

High because of community or other

Significant No loss of life expected, but the possibility recognised. No urban

significant developments

development, and no more than a

downstream.

small number of habitable

Loss of identifiable life expected

Low No loss of life expected.

structures downstream. Excessive economic loss such as serious damage to communities, industrial, commercial or

Appreciable economic loss, such

Minimal economic loss, such

as damage to limited land areas,

as farm buildings; limited damage to agricultural land,

agricultural land or facilities, important utilities, the dam itself or

relatively important public utilities, the dam itself or other

other storages downstream.

storages downstream.

secondary roads, minor railways,

Repairs to dam not practicable. Dam Repairs to dam practicable or essential for services.

alternative sources of water/power

minor roads etc.

Repairs to dam practicable. Indirect losses not significant.

supply available. Npte that items refer to incremental losses and effects due to dam failure. Hence, an

assessment of the damage and loss of life due to the earthquake with and without breaching of the dam may be made.


19

4.2 Selection of the Design

Earthquake 4.2.1 General Approach There are two general approaches to selection of the "design earthquake".

(a) The Deterministic, Prescribed Approach

earthquake either on an identified fault, or f an AEP vs ground motion graph determinec the seismotechnic province provided for dam. For high hazard dams, the MCE ma; adopted to determine the MDE, but requires a knowledge of the fault(s) on w] the MCE may occur. This is impracticable in Australia. Table 4 summarizes the practice of a numbs organisations.

This is the method which has been most widely used to date. The design earthquake load (MDE or OBE) is determined by consideration of the hazard category of the dam, and is selected as an earthquake with a given AEP. For example, a significant hazard category dam

might be designed for a 1 in 1000 AEP

Table 4. Methods Used by Dam Authorities to Estimate Maximum Design Earthquake (MDE) Organisation

Method

Reference

ICOLD

a) Normally equal to MCE obtained deterministically,or "50% probability or higher of not being exceeded in a large

ICOLD (1989)

number of years.

b) If no hazard to life (ie. low hazard dam), use less than MCE based on economics.

Generally equated to controllingMCE.

FEMA (1985)

US Bureau of Reclamation

MCE, or where there are no specific faults,

USER (1989)

(USER)

an AEP of 1 in50,000(1).

National Research Council (of

MCE for high hazard dams and less than MCE for low hazard dams.

National Researc

MCE for high hazard, 0.5 MCE for significant hazard, 1 in 200 AEP for low

Salmon and von

Federal Emergency Management

Agency (of USA) (FEMA)

USA) EC Hydro, Canada pre 1993

h!

Council (1985)

(1993)

hazard.

EC Hydro, Canada post 1993 Canadian Dam Safety

Risk based approach, see Notes(2).

EC Hydro (1993]

MCE or 1 in 5000 to 1 in 10,000 AEP for

CDSA (1994)

Association (CDSA)

very high consequence; 0.75 to 1.0 MCE, or

1 in 1000 to 1 in 5000 AEP for high consequence, 1 in 100 to 1 in 1000 AEP for low consequence, see Notes(3).

New South Wales Dams Safety Committee (Australia

For high hazard the most severe seismic, loading that could reasonably occur in the area; for significant hazard dams AEP 1 in 1000, or base on risk analysis.

NOTES:

assume they refer to high hazard

(1) USER do not specify hazard

dams.

category, but it is reasonable to

20

NSWDSC(1993

AN COLD Guidelines for Design of Dams for Earthquake

(2) For dams with greater than 100

earthquake is low (based on historic data).

potential premature deaths,

This appears to be recognised to some extent

maximum tolerable AEP is 1 in 100,000 and MCE should be

at least in the USBR (1989)

probability of dam failure with

recommendation of an AEP for MDE of 1 in 50,000 for high hazard category dams, although they do not justify the figure in

potential to cause premature fatalities

these terms. It also appears to be recognised

should be less than that resulting in 0.001 fatalities per year of dam

in the recommended AEPs in Canadian Dam

adopted. For other dams, the annual

Safety Association(1994)

operation, except that no identified

individual should have more than 0.0001 probability of fatality per year of dam operation.

(3) The boundary between very high and high consequence categories is an

incremental loss of life of 100 persons. Low consequence is,

roughly, equivalent to significant hazard, very low to low hazard in Table 3.

(b) The Risk Based Approach In this method, the adequacy of the dam is considered by calculating the probability of breaching of the dam and the expected incremental loss of life due to the dam breaching. The acceptability of the dam is then tested by comparison of the calculated values with societal and individual risk criteria. In this approach, the concept of MDE is not applicable, since one must consider the risk arising from the full range of possible earthquake events.

There are two problems with each of these methods:

• the risk based approach is a relatively new one, and the methods for assessing the

conditional probabilities of breaching are still being developed, and can be time consuming and costly. It is also difficult to justify only considering the risk due to earthquake, since logically one should sum

the risk from flood, earthquake and other causes before comparing with acceptable

risk criteria. This means that studies should consider flood and other causes, increasing

the complexity and cost of the study. There is some evidence of a trend towards a risk

based approach to earthquake design. The most detailed discussion of this is given in Salmon and von Hehn (1993) who describe the BC Hydro approach at that time. BC Hydro (1993) give more details of these methods, which are being used by that organisation. Finn (1993) indicates that in California there is a gradual shift from use of MCE to a Safety Evaluation Earthquake (SEE). He indicates that in selecting this, consideration is given both to the probability of occurrence and the consequences of failure.

(c) Recommended Approach • the deterministic/prescribed approach fails to account for the fact that the conditional probability of breaching of the dam, given that the MDE (or other selected earthquake) occurs, is not 1, and may be very different for different types of dams and foundation conditions. Hence, the adoption of a probability of the initiating event, ie. the / earthquake, may result in a very wide range of probabilities of dam breaching. This is quite different to the case for flooding and embankment dams, because the probability that a dam fails on overtopping is high, but the probability that a dam breaches due to an

Based on consideration of all of the factors, the Working Group has concluded that for high and significant hazard dams, the design for earthquake should where practicable be based on risk assessment methods. These are detailed in section 4.2.2. For low hazard category dams,

a deterministic approach using , a MDE equivalent to an AEP of 1 in 200 may be adopted, but the decision will probably be based on economic considerations.

It is realised that many will be more comfortable with an AEP for the MDE related


21

directly to the hazard category of the dam. The information in Table 4 may be used if this

concrete gravity dams, overturning, j sliding— see section 4.4.2 (PB)

approach is followed. In particular, reference

should be made to the Canadian Dam Safety Association (1994) criteria which are

(iii) assess the probability of failure for ej range of ground motion by multiplying)

reproduced in Appendix C. However, it must

AEP with PBc, ie. PB = PE x PBC ;

be recognised that there is an implied probability of breaching and probability of loss of life in these values. In reality these probabilities are very variable, eg. possibly 1 to 3 orders of magnitude different for PEB (the probability of breaching) of a homogeneous earthfill dam compared to a concrete gravity dam. The potential for loss of life in the event of breaching is also very site dependent.

TableS (iv) sum the probabilities to give the ovl annual probability of failure due earthquake. To this, should be added th annual probability of breaching due flooding, and other causes (eg. pip instability) making allowance ! dependence, independence and mu

exclusivity of the events. The Work Group recognises, however, that in m

Where causative faults are identified in the vicinity of the dam, the MDE from these faults should be determined. For very high hazard category dams, the more critical of the ground motion obtained from the MCE on these faults, and probabilistic approaches should be adopted. For significant and high hazard category dams, an AEP should be assigned to the CMCE determined from geological and geomorphological assessment of the causative

faults, and the MDE obtained this way compared to that obtained from the probabilistic approach, with the larger being adopted. 4.2.2 Risk Assessment Method for Design for Earthquake

cases, these studies will be d independently, or the risk from earthqu may far exceed those due to flood or of causes, and the probability of failure dw earthquake alone will be carried forwan the next step.

(v) determine the incremental loss of expected for the dam breach using method detailed in section 4.2.5. It sho be noted that there may be differ estimates of the loss of life for different earthquake ranges, and earthquake, flood and other causes (vi) determine the average individual risk

the Population at Risk (PAR) from (v) i The steps involved in this process are: (i) determine the AEP of earthquake ground motion (PE) over the range of earthquake events which may affect the dam. Table 5 gives an example for AEP vs ground acceleration

(ii) determine the conditional probability (Pbc) that for each of the ground motion ranges (eg. 0.125g to 0.175g in Table 5) the dam will breach. In assessing this conditional probability all modes of failure should be considered and the probabilities combined, making allowance for interdependence and mutual exclusivity or otherwise (eg. for embankment dams, slope instability, piping, liquefaction/instability, and for

22

the PAR, and the risk to the individ most at risk (vii) assess the acceptability of risk using criteria detailed in section 4.2.3.

The total risk to property is calculated ii similar way, except that PBC is replaced b; vulnerability term, ie. the proportion of element at risk, eg. house, bridge, which is 1 as a result of the earthquake (Vbc)- Vbc is in range 0 for zero damage to 1.0 for total loi Then total risk = 2 PE x PBC x (value of elemf summed over the range of earthquake ev< and the elements at risk.


TableS. Acceleration

Annual

Probability <0.075g 0.075gto 0.125g 0.125gto 0.175g 0.175g to 0.225g 0.225g to 0.39 >0.3g

TOTAL

0.874 0.100 0.015 0.007 0.003 0.001 1.000

Conditional(1)

P (2) rB

Probability of (Pbc) 0.0005 0.005 0.05

0.0004 0.0005 0.0007 0.0007 0.0009 0.0005 0.0037

0.1 0.3 0.5

(1) Given the earthquake occurs (2) PB = annual probability x PBC being a member of the public, ie. the individual most at risk, to a risk greater

4.2.3 Acceptable Risk

than 10"5 per annum.

The calculated risks are compared to the acceptable risks. What is acceptable will

For existing dams, individual risks up to 10 times those for new dams could be tolerable subject to application of the ALARP principle.

depend on many factors, including social,

political and economic, and the decision may be made by persons other than the person carrying out the risk assessment.

It is recommended that the calculated risk be related to the acceptable risks outlined in the ANCOLD Guidelines on Risk Assessment (ANCOLD, 1994), which is due for revision in

1998. The relevant ANCOLD (1994) recommendationsare:

Dam owners are to assess the

acceptable risks for workers at and below their dam sites.

ANCOLD (1994,1996), recommends that a dam should comply with both individual risk and societal risk criteria.

Societal Risk:

ANCOLD (1994) also indicates that in risk

Ensure that new dams, and dams being

upgraded, satisfy the Societal Risk criterion given by the Objective curve of Figure 6. Ensure that existing dams satisfy the Societal Risk criterion given by the Limit curve of Figure 6, but carefully

consider the ALARP (As Low As Reasonably Practicable) principle. Individual Risk: ¦/ For new dams and upgrading of existing dams, ensure that the average

risk of death to particular members of the public from dam failure does not exceed 10"6 per exposed person per annum. Do not subject any person,

assessment, if the cost per life saved is less than

A$500,000 (1990 values) the investment can be considered as certainly worthwhile. Appendix D presents some additional information relating to acceptable and actual risks.

As discussed above, the probability which should be used in Figure 6 is for all modes of failure - earthquake plus flood plus other causes. In the event that only earthquake is being considered, allowance should be made for flood and other causes, eg. one might assign

l/3rd of the acceptable probability of failure (breaching) to earthquake, but consideration should be given to the relative probability of failure under flood, earthquake and other static loading.


23

10 104 10J N, number of fatalities due to dam failure

Figure 6. ANCOLD amended interim societal risk criteria (ANCOLD, 1996). 4.2.4 Estimation of the Probability of Dam Breaching from Flood, Earthquake and Other Causes (a) Worldwide Experience of Behaviour of Dams It is common to assume that embankment dams

will fail when they are overtopped by flooding; and concrete gravity dams will fail when the IFF level is reached. The IFF is based on assumed uplift pressures, foundation strengths, etc, or on the flood level of record. In both cases, there is an implicit assumption that the probability of breaching PB is 1.0. Many embankment dams, particularly those with steel

withstand some overtopping without' Many concrete gravity dams won

flood levels higher than the IFF dep how conservativelythe IFF had been Hence, in reality, PB is less than 1.0 f< Table 6 shows the detail of reasons of embankment dams, other than appurtenant works (which indue overtopping).

mesh reinforcement of the downstream zone

(for floods during construction), would

24


Table 6. Most Frequent Classification of Deterioration and Dam Failure in Embankment Dams

(ICOLD, 1983). Classification

Percentage of Total

Classification Deterioration Failure Percolation (foundation) Percolation (dam) Slope protection Differential movement Internal erosion (dam) Downstream slips Internal erosion (foundation) Compaction Upstream slips Bonding between concrete structures and embankment

Deformation and land subsidence Watertight systems Earthquakes Drainage system and filters Shear strength

27 25 19 17 17 11 11 8 8 7 5 4 4 4 4

Ratio of Failures to Deterioration As a Percentage

26 38

9

<1

NA

31 49 16 17

18 29 15 15 14 6 38

12 5 6

15

7

14

<1 <1 <1

NA NA NA

5

13

NOTES: (1) NA = Not available. (2) 'Percolation' is often related to lack of seepage control measures. (3) Deformation is often associated with percolation, internal erosion, shear strength. Land subsidence is a minor cause.

(4) Total number of dams 9900 Total number of deteriorations (a) 432 (4.4%) Total number of failures (a) 43 (0.43%) (a) Not including appurtenant works, which total 39 failures. These account for most overtopping failures due to inadequate spillways.

As discussed ICOLD (1983) and reproduced in Fell, MacGregor and Stapledon (1992), less than 1% of dam failures are due to earthquake, and approximately 50% of failures are due to causes other than flood overtopping and earthquake.

failure can be calculated to be as shown in Table?. Similar calculations have been done to produce Tables 8 and 9 for concrete dams (excluding masonry dams), and Table 10 for appurtenant works.

Given that 0.83% of all embankment dams

surveyed by ICOLD failed, the probability of

/


25

I

:

Table 7. Average Probability of Failure of Embankment Dams Over the Life of the Dam,

on ICOLD (1983). Probability

% of Failures

Cause of Failure

of Failure < Flood overtopping and failure of other appurtenant works

Slope instability Internal erosion and poorly controlled seepage in embankment Internal erosion and poorly controlled seepage in foundation

Other

48% 8% 28% 12%

1 in 250 ! 1 in 1500 1 in 425 |

4%

1 in 3000 ;

1 in 1000 1

100%

Total

1 in 120 ]

NOTE: (1) Combined probability of internal erosion failure «1 in 300 Table 8. Most Frequent Classification of Deterioration and Dam Failure in Concrete Di

(ICOLD, 1983). Ratio of Failures to

Percentage of Total Classification Deterioration Failure

Classification Inadequacy of site investigation Shear strength foundations Percolation foundation

Internal erosion (foundation) Grout curtain etc Drainage in foundation

Uplift affecting foundation Shape of dam and valley (arch) Shape of dam and valley (gravity) Tensile stresses in dam

7 9 28 21 11 11 12

4 5 11

Deterioration As a Percentage

14

60 43

14 5 29 10 5 5 5 10 5

5 37 25 12 11 33 40

9

NOTES: (1) Total number of dams 4250 Total number of deteriorations (a) 251 (5.9%) Total number of failures (a) 12 (0.3%) (a) Not including appurtenant works, which total 8 failures (2) Does not include masonry dams.

Table 9. Most Frequent Classification of Deterioration and Dam Failure in Concrete Dams (

the Life of the Dam (ICOLD, 1983). Cause of Failure

% of Failures

Probability of Failure

F oundation strength

40% 15%

Foundation erosion and seepage

25%

Foundation uplift and drainage

5% 3% 12%

Flood overtopping and failure of other appurtenant works

Tensile stresses in dam

Other Total

100%

NOTE: (1) Combined probability of foundation failure «1 in 475

26 ANCOLD Guidelines for Design of Dams for Earthquake

1 in 525 1 in 1400 1 in 850 1 in 4000 1 in 7000 1 in 1750 1 in 210

Table 10. Most Frequent Classification of Deterioration and Dam Failure in Appurtenant Works

(ICOLD, 1983). Ratio of Failures to

Percentage of Total

Classification

Classification Failure

Deterioration

Deterioration Percolation information

6

Internal erosion of foundation

8

13 11

Removal of rip-rap

<1

4

Corrosion of steel and other materials

3 10 24 10 14 11 12 9

4

Structural failure Excessive flow, including spillways

Solid materials carried by waterflow Local scouring Malfunction of discharge equipment Erosion by abrasion (concrete) Erosion by cavitation

As a Percentage

29 19 40 18 10 44 8 11 13 NA NA

7 73 5 11

9 <1 <1

The ICOLD (1983) study also assembled data

These tables are for average dams, and the

figures in Tables 7 and 9 are over the life of the

on the time of deterioration after construction.

dam. The average age of the dams in the ICOLD survey was -30 years, so average

This data is shown in Table 11.

annual probabilities of failure can be obtained by dividing the figures in Tables 7 and 9 by 30. This gives the average annual probability of

It will be noted that a disproportionate number of failures occur on first filling. This can be taken account of in the overall assessment, eg. for embankment dams, 40% of failures occur

failure of embankment dams as Ô.00015, or 1 in 6000, and of concrete dams «0.0001, or 1 in

10,000 (both figures excluding overtopping and

during first filling, so the average probability of failure during first filling is 0.4 x 0.43% or «1

other failures of appurtenant works).

in 500. The equivalent value for concrete dams

is also«l in 500. Table 11. Distribution in Percentage of Deterioration an Failure Cases in Terms of Time After Construction.

Failure Time

Concrete Dams

Failure Construction

First filling First 5 years After 5 years Not available

11 67 6 11 5

These figures do not take account of the design of the dam, the quality of investigation and construction, or the fact that dam engineering has improved since the ICOLD survey in 1975,

Embankment Dams

Deterioration

Failure

Deterioration

7 11 10 30 42

14 40 11 30 5

11 25 20 26 18

For example, a homogeneous earth dam

Logically, some types of dam are more

constructed of dispersive soil, and showing signs of excessive seepage, is more likely to fail and breach than a central core earth and rockfill dam with well designed and constructed filters. McCann et al (1985) have made an attempt at quantifying this, but their study does not

susceptible to failure and breaching than others.

account for zoning or condition of the dam.

or the degree of monitoring and surveillance.


27

1

A study by Hackney (1994) of 65 of the case histories of embankment dams (including 15 failures)on which the ICOLD survey was based, showed that homogeneous earthfiII and

zoned earthfilldams which have no filter & were disproportionately represented in incidents. Figure? shows the results.

Figure 7. Dam category vs percentage of total observed incidents (Hackney, 1994). Category CI- Homogeneous Earthfill } Category C2- Earthfill with Toe Drain } No filters Category C3- Zoned Earthfill } Category C4 - Earthfill with Horizontal Drain Category C5 - Earthfill with Vertical and Horizontal Drain Category C6 - Earth and Rockfill - Central Core Category C7 - Earth and Rockfill - Sloping Upstream Core Category C8 - Concrete Face Rockfill Category C9 - Earthfill with Concrete Core Wall

The sample of dams studied by Hackney (1994)

more susceptible to failure and accident

was reasonably representative of the ICOLD (1983) study population of dams in respect to

others.

age of dam and dam height. However, there is

The discussion so far has referred to dams ui non earthquake conditions. The performam dams under earthquake conditions has genei

not sufficient data in the ICOLD study to determine the total number of dams in each category, so it is not clear whether the reason for the overrepresentation of homogeneous

earthfill and zoned earthfill dams in the incidents in Figure 7 is due to a higher

been good with few dams suffering m damage except where liquefaction has occii in the dam or its foundation. Seed (19

ICOLD (1986), USCOLD (1992), NSW1

proportion of those dams being present, or

(1993) and Hinks and Gosschalk (1993)

whether they are more susceptible to failure and accidents. Given that a large proportion of the incidents are due to internal erosion of the dam

some details.

and the foundation, and there is poor control over these in homogeneous and zoned earthfill

dams, it seems likely that these types of dam are

However, this needs to be considered in

context that relatively few dams 1 experienced major earthquake loading, so too much comfort should be taken in performance.

28


What can be expected in major earthquake is:

are described in ANCOLD (1994). The method requires the use of judgemental

• materials susceptible to liquefaction may

"expert" opinion on assessing probabilities, and is in use in Canada, eg.

lose a significant amount of strength, and large deformations may occur. This is

discussed at length in Section 5

Salmon (1995), Salmon and Hartford

(1995 (a),(b)) (and more recently in

• embankment dams can be expected to settle

Australia, eg. Prospect Dam,

and crack. Most cracking is likely to be

(Landon-Jones et al, 1995).Expert opinion can be supported by reliability and other analyses that help to describe the response of the dam system to earthquake loading. It is recommended that where practicable

longitudinal, but transverse cracking may occur

• concrete dams can be expected to crack and may move on their foundations. This may

significantly affect the available shear strength on the foundation, foundation uplift

event tree analysis is carried out, since this

pressures, uplift pressures and tensile

design, construction, monitoring and surveillance of the dam.

strengths in the dam Hence, it is logical that the dams will be more susceptible to failure after experiencing the earthquake, than before. The extent to which

this affects the probability of failure and breaching depends on:

includes assessment of the details of the

(ii) Empirical approach. An approximate empirical method for estimating the probability of failure has been proposed in Fell (1995). As pointed out in the paper, the method is based on many assumptions

• the severity of the earthquake and damage to

which cannot be backed up with reliable data, and it should only be used for

the dam, for example, homogeneous earthfill

preliminary assessments of existing and

dams will have a higher probability of

new dams.

failure post earthquake cracking, than central

core earth and rockfill dam, with well designed filters and free draining rockfill; a concrete gravity dam on strain weakening interbedded shale and sandstone, is more

likely to experience stability problems than one on granite with favourably oriented

joints • the detailed design of the dam • the water level at the time of the earthquake and how quickly it can be lowered • the potential for earthquake aftershocks. At this time no specific guidance can be given as to how these issues can be quantified.

(b) Recommended Method There are two approaches which can be taken:

(i)^ Event tree analysis, where the potential mechanisms leading to failure and breaching are identified, and probabilities assigned to these events, leading to an

overall assessment of the probability of breaching. The procedures for this method

4.2.5 Estimation of the Loss of Lives if a Dam Breaches

(a) General The most comprehensive method for estimating the loss of life from a dam breaching has been to carry out a dam break analysis and use the

USER (1989) "Policy and Procedures for Dam Safety Modification and Decision Making" approach to estimate loss of life, given the flood velocities, depths and other data. However, the

USER procedure has drawbacks, notably the dichotomous nature of the critical warning time parameter.

The paper "Predicting Loss of Life in Cases of Dam Failure and Flash Flood", by DeKay and McClelland (1993), provides an improved method for estimating the potential loss of life due to severe flooding. It is derived from the historical record of dam failures and flash flood cases via logistic regression. This uses the data

on which the USER (1989) method is based, as well as some other cases, and analyses the data


29

The number of residents iq

somewhat more rigorously than the USBR method. Because of this, and because it gives a single continuous equation, the DeKay and

inundated area may be estimati

one of the following methods:

McClelland (1993) method is recommended. A summary of the method is included in these guidelines, but it is important that users read that paper in detail, because it explains the empirical basis on which the method is

1. Use the current cens

population of the town| region only if it

determined. In particular, users should be

completely inundated. 2. Use census data based

aware of the wide confidence intervals applying

postal codes within

to the regression equations, as evidenced by the

inundated area.

differences between the actual and predicted loss of life given in Table 1 of the paper.

(b) The DeKay and McClelland Method

3. Multiply the number homes in the area by average number of peoj per home for that particu|

The general equation for potential loss of life is:

4. Conduct a house -to-hou

area.

survey.

LOL=

PAR 044\ (0 759(WT)-3 790(For
1 + 13.277(PAR )

The assessment process needs

WT

Potential loss of life Population at risk Warning time (hours)

Force

Forcefulnessof flood waters

where LOL

PAR

subdivide the numbers dependini distance (and, hence, warning t)

)

from the dam.

Once the number of residents infl area is estimated, the population atj is found by multiplying the numbe

1 for high force 0 for low force

residents by an annual exposure fac

This is the fraction of a year tha| typical individual would spend home. A typical exposure time? residents is in the range of 0.6 to 0.8|

The method is only statistically rated for a population at risk of less than 100,000, and is not applicable to cases of dam failure equivalent to Vaiont Dam, where a massive landslide into

The population at risk for facilil other than residence (such as schod industrial buildings, parks, shoppj centres, etc) should be found multiplying the estimated aver! number of people within the facilit

the reservoir resulted in 1209 deaths from one village of 1345 persons alone. In the DeKay and McClelland method the following assumptions are made:

an exposure factor, which may refl

(i) Population at Risk (PAR) is defined

the fraction of the year that the facil is open. If the number of people andj exposure factor cannot accurately! found, then estimates should be ittf using the best overall judgemetl However, it should be noted thai

as the population at risk of "getting their feet wet" if they took no action to evacuate, ie. regardless of the depth or velocity of the flood water. No-one

who is more than three hours flood travel time below the dam should be included in the PAR. The PAR will usually be calculated from a dam break analysis to determine the area which will be inundated.

30

'I

some areas these "special" populati

can be very large, and need toj estimated accurately.

(ii)

Warning Time (WT). There is single method for estimating war!


time. In their study database DeKay

and McClelland (1993) used the recorded WT (adjusted as described in the paper). USBR( 1989) states that:

determined on a case by case basis. For unmanned dams in remote areas, warning times may be less than the flood arrival time, because the failure may go unnoticed.

"Specifically, the time to be estimated is the time at which the public warning process has been

officially set in motion and the first individuals for each PAR are being

(iii) Force. DeKay and McClelland (1993) describe Force as flooding lethality indicated by variations in the depth and velocity of the flood waters. In the

warned to evacuate. Again, this time is expressed as the number of

general equation Force is treated as a

hours prior to the arrival of flooding at the location of the

one for high force and zero for low

dichotomous variable with values of force.

PAR". In other words, they state that warning

time is equal to the difference between the time the public evacuation warning is initiated and the actual flood arrival time for the population at risk. Because of the difficulty in estimating the time that the public evacuation warning would be initiated, a

The term high force refers to flood waters that are likely to be deep and swift. This would be typical in a narrow valley or canyon area.

Substituting a Force value of one into the general equation gives the following expression for potential loss of life in a high force or canyon area: LOL=

PAR

conservative estimate for dams which

have full time operating staff would be to use the time at the beginning of dam breach (provided the operator can issue a warning). That is, warning time is

equal to the flood arrival time as shown on the inundation maps. According to

DeKay and McClelland (1993): "Such assumptions may be reasonable for earthquake

-induced failures (and perhaps terrorist acts), but other modes of

failure will typically allow for longer WTs and greater evacuation benefits. We assume

that the case in which an

1 +13211 {PAR0 44 )g(2'982(',T)"3 79())

The term "low force" refers to flood

waters that are likely to be shallow and slow. This would be typical in the mature reaches of a river where there

are wide alluvial flood plains. With a Force value equal to zero, the general

equation gives the following expression for potential loss of life in a low force or plain area:

LOL=

PAR 1 + 17)211 {PAR0 u)e^159i}VT))

evacuation is ordered at the time of dam failure represents the worst-case scenario and the most reasonable baseline for assessing

For cases where the population at risk is located partly in a narrow valley area and partly in a plain area, it may be

the benefits of additional WT".

necessary to divide the PAR. However,

DeKay and McClelland (1993) point The actual warning time will depend on the nature of the dam failure, whether there is surveillance of the dam, flood preparedness plans, and will have to be

out that because the loss-of-life

equations vary nonlinearly with PAR and WT, this may lead to an overestimation of the potential loss of


31

life. They indicate that division of the total population at risk into more and more groups can easily lead to significant overestimates of the

potential loss of life. DeKay and

McCleIland(1993) suggest that: "the PAR should only be divided when there are likely to be significant differences in the forcefulness of the flood waters reaching the PAR within 3 hr of

considered when considering eco losses, and social, cultural; environmental losses which;

attributable to the dam failure earthquakes, there may be ext| property damage and loss oj independent of dam failure.

4.3 Selection of the Operating Ba

Earthquake (OBE)

the dam. In general, no more than

The OBE should relate to the life o

two subpopulations should be used in predicting potential LOL".

structure, the economic and i consequences of damage to the dam a1

Therefore, in applying these guidelines, it is recommended that the population at risk may be divided into two groups, but only when there is a significant difference in forcefulness or warning

time. Any situations that may call for a greater division of the population at risk should be considered for a more

in-depth study of the potential loss of life.

likely to be different for different parts o dam. Typical values are in the range AE 200 to 1 in 1000. For many situation^ requirements of Australian Standard AS1 f Minimum Design Loads on Structures, P*

Earthquake Loads, will apply. The accele coefficients for AS 1170.4 are based on ai chance of exceedance in 50 years (equival an AEP of 1 in 475). These are mo depending on occupancy type (importanc the structure and likely consequence failure).

(iv) Incremental Consequences of Failure. In earthquakes, there may be

extensive property damage and loss of life independent of the dam failure. Therefore, when assessing the risk due

to a possible dam failure event, it is

It should be noted that care must be take selecting OBE for appurtenant structures, w on first assessment, may not appear critic the dam. In many cases, the ability o appurtenant structures to operate success

same extreme event. In equation form,

after an earthquake, may be critical to whe the dam may breach. Eg. if spillway i become inoperable, delayed failure \ overtopping may result; inability to emp reservoir behind a badly damaged dam greatly increase the probability of J earthquake breaching; a dam may be critic^

this statement would look like:

water supply, so outlet works must contin

necessary to use the incremental consequence of failure. The incremental loss of life for an event is

equal to the losses that would occur with dam failure minus the losses that would occur without dam failure for the

operate, or be able to be repaired rapidly

INCREMENTAL LOL =

an earthquake.

LOL WITH DAM FAILURE ~ LOL WITHOUT DAM FAILURE

For some cases, the incremental loss of

life may be equal to the loss of life with dam failure. This, however, should be

decided when the actual procedures are being applied to a risk assessment for a specific dam. This also needs to be

32

4.4 Concurrent Load Combinatio The selection of the water level in the da the time of the earthquake should be detern taking into account the combined probabi of the water level (related to floods and sea variations in the water level in the dam), an earthquake. This will preclude, or


correctly render very unlikely, the application

concurrently of low AEP floods with low AEP

4.6 Response Spectra and Accelerograms

earthquakes.

The joint probability of other unusual load conditions, eg. high pore pressures soon after construction or raising of a dam needs to be considered.

It is also essential to consider likely water levels in the dam after the earthquake. It may, for example, take some months to draw down the

water level in the reservoir behind a badly damaged dam, in which time additional flooding may occur, leading to higher water levels and greater chances of breaching of the dam.

In some circumstances, particularly for the analysis of concrete dams, and dynamic analysis of earth and rockfill dams, response spectra, and accelerograms suited to the site and

the design earthquakes will be needed. This is discussed in section 7.4. Advice from

seismologists should be sought when selecting suitable response spectra and accelerograms.

5. DESIGN OF

EMBANKMENT DAMS AND ANALYSIS OF LIQUEFACTION

4.5 Earthquakes Induced by the

5.1 Effect of Earthquake on

Reservoir

Embankment Dams

As discussed in Section 2.7 and in ICOLD

Earthquakes impose additional loads on

(1983) and ICOLD (1989), there is documented

embankment dams over those experienced

evidence to prove that impounding of a

under static conditions. The earthquake loading

reservoir sometimes results in an increase of earthquake activity at or near the reservoir.

is of short duration, cyclic and involves motion

ICOLD (1983) conclude that:

in the horizontal and vertical directions. Earthquakes can affect embankment dams by causing any of the following:

- earthquakes of magnitude 5 to 6.5 were

- settlement and cracking of the embankment,

induced in 11 of 64 recorded events - the greatest seismic events have been

associated with very large reservoirs (but there is insufficient data to show any definite correlation between reservoir size and depth

and seismic activity) - the load of the reservoir is not the significant factor, rather it is the increased pore water

pressure in faults, leading to a reduction in shear strength over already stressed faults.

- in view of the above, a study of possible induced seismic activity should be made at least in cases where the reservoir exceeds

109m3 in volume, or 100m in depth

ICOLD (1983 and 1989) give more details and references on this issue. /

overtopping of the dam - instability of the upstream and or downstream slopes of the dam - differential movement between the

embankment, abutments and spillway structures, increasing the likelihood of leakage and piping failure - liquefaction or loss of shear strength of saturated granular soils in the embankment or its foundations due to increase in pore

pressures induced by the cyclic loading of the earthquake - differential movements or instability on faults or other low strength seams or defects

Reservoir induced seismicity has been recorded in Australia.

particularly near the crest of the dam - reduction of freeboard due to settlement, which may, in the worst case, result in

passing through the dam foundation - overtopping of the dam in the event of large tectonic movement in the reservoir basin, or

by seiches induced upstream


33

- overtopping of the dam by waves due to

earthquake induced landslides into the reservoir from the valley sides - damage to outlet works passing through the

embankment leading to leakage and potential piping erosion of the embankment. The potential for such problems depend on: - the seismicity of the area in which the dam is sited, and the assessed design earthquake - local foundation and topographic conditions - the type and detailed construction of the dam - the water level in the dam at the time of the earthquake.

analysing this type of problem - rati we should concentrate our efforts

those dams likely to present problei either because of strong shaki involving accelerations well in excess 0.2g, or because they incorporate lar

bodies of cohesionless materii (usually sands) which, if saturated, m lose most of their strength duri earthquake shaking and, thereby, lead undesirable movements".

In other words, it is only the potent liquefaction which is critical for such dam

additional construction measures (over those

It is important to consider what is me "well built dams" in applying this j

needed for static conditions) will depend on

statement.

The amount of site investigation, design, and

these factors, the hazard category and whether the dam is existing or new. There are four main issues to consider:

• the general (or "defensive") design of the dam, particularly the provision of filters, to

Since one of the main effects of earthqual induce cracking, this will increas likelihood of piping failure, so earth, am and rockflll dams which are built withoi designed filters may not be considei

prevent or control internal erosion of the dam and the foundation, provision of zones

"well-built". Similarly, care should be ta

with good drainage capacity (eg. free

older puddled core, and concrete con dams, which have poor erosion control, a

draining rockflll) • the stability of the embankment during and immediately after the earthquake • deformations induced by the earthquake (settlement, cracking) and dam freeboard • the potential for liquefaction of saturated sandy and silty soils in the foundation, and possibly in the embankment, and how this affects stability and deformations during and immediately after the earthquake.

considering as "well built" some of Aus

usually poorly compacted.

5.2 General ("Defensive") Desig Principles for Embankment Dams

ICOLD (1986), Sherard (1967), Seed | and Finn (1993) present lists of dei design measures. The general philosoph apply logical, commonsense measures

These issues are considered in the following

design of the dam, to take account <

sections, with references which give more

detailed information.

cracking, settlement, and displacements may occur as the result of an earthquake, measures are at least as important (pr<

It should be noted that many low hazard category dams may not warrant extensive

more so) as attempting to calculate acci the stability during earthquake, or the

investigation and design measures. ICOLD

deformations. The most important me

(1986) quote Seed(1979):

which can be taken are:

"Since there is ample field evidence that well-built dams can withstand moderate shaking with peak accelerations up to at least 0.2g, with no harmful effects, we should not waste our time and money

34

(a) Provide ample freeboard, above i operating levels, to allow for sett or slumping or fault movements;

displace the crest. For exampl might adopt a narrow spillwa] large flood rise (and, hence,


freeboard) instead of a wide spillway with small flood rise and thus usually a

and if the core is a highly dispersive soil, it could be modified with lime to

lower freeboard, provided the costs were similar.

make it non dispersive. A related issue

(b) Use well designed and constructed filters downstream of the earthfill core (and correctly graded rockfill zones downstream of a concrete face for

concrete face rockfill dams) to control erosion in the event the core is cracked in the earthquake. For larger dams, full

is the detailing of the crest to control internal erosion in the event of settlement and cracking in an earthquake. For example, it will be advantageous to take filters to the crest level, not just to within say one metre of the crest. Also, the common practice of using steeper slopes near the crest

(usually only to provide for camber of

width filters (2.5m to 3m) might be

the crest), and the use of wave walls in

adopted instead of narrower (1.5m say) filters placed by spreader boxes. This would give greater security in the event

concrete face rockfill dams, will make the crest more susceptible to damage

(Sherard, 1967).

of large crest deformations.

(g) Flare the embankment core at abutment (c) Provide ample drainage zones to allow for discharge of flow through possible cracks in the core. For example, ensure

contacts, where cracking can be expected, in order to provide longer seepage paths. Just as (or more)

that at least part of the rockfill is free draining, or that extra discharge capacity is provided in the vertical and horizontal drains for an earthfill dam

provision of filters downstream of the contacts. This detailing is discussed in

with such drains.

Fell etal (1992).

(d) Avoid, densify, drain (to be

important, is to consider the detailing of the contact with concrete walls, and the

(h) Locate the core to minimize the degree

non-saturated) or remove potentially

of saturation of materials (eg. use

liquefiable materials in the foundation

sloping upstream core). (Finn (1993) also suggests positioning chimney

or in the embankment. There are a number of other measures which are

drains near the central section of the embankment). These measures are

listed by Sherard (1967), Seed (1979) and Finn

intended to reduce to a minimum the

(1993). These include:

extent of saturated zones which are

(e) Use a well graded filter zone upstream

more likely to reduce strength on cyclic loading. It is particularly relevant

of the core to act as a crack stopper,

possibly only to be applied in the upper

where sand, silty sand and sand-gravel soils are present as these are most

part of the dam.. The concept is that in

susceptible to liquefaction.

the event that major cracking of the core occurs in an earthquake this filter material will wash into the cracks, limiting flow, and preventing enlargement of the crack. If well designed filters are provided downstream, this upstream filter is of / secondary importance.

(i) Stabilize slopes around the reservoir rim (and appurtenant structures, such as

spillways) to prevent slides into the reservoir.

(j) Provide special details if there is danger of movement along faults or seams in

the foundation. (f) Provide crest details which will minimise erosion in the event of overtopping, eg. by having a wide crest,

(k) Site the dam on a rock foundation, rather than soil foundation (particularly


35

if it is potentially liquefiable) where the option is available.

(1) Use well graded (densely compacted) sand/gravel/fmes, or highly plastic clay

Assuming that for areas with earthquakes magnitude 6.5 or le maximum ground acceleration! than 0.3g, the slopes are 1.35 horizontal:! vertical,then

for the core, rather than clay of low

plasticity (if the option is available) (Sherard, 1967).

- for design earthquakes magi 6.5 to 7.5, and ground accelera

up to 0.5g, slopes should flattened to about 1.65H:IV

When assessing an existing dam, the use of these "defensive design" measures is seldom

- for design earthquakes magnii

practical (except in remedial works). However,

7.5 to 8.5, with gfj

it is useful to gauge the degree of security the existing dam presents by comparing it with this list. Where the dam fails to meet many or most of these features, particularly (a) to (d), this may be a better guide than a lot of analysis, to the fact that the dam may not be very secure against

accelerations up to 0.5g, si

earthquake.

should be flattened to 1.8H:IV.

They suggest these are only general guidel and that these flatter slopes may only nei apply to the upper part of the dam accelerations are the greatest. As pointej

Apart from the defensive design measures, there are some other general points which can be made:

by Finn (1993) this general philosop! defensive design may be applied to other of dams. However, the use of such

slopes has not been widely adopted. (m) Some dam zonings are inherently more earthquake resistant than other embankment dam types. In general, the

following would be in order of decreasing resistance: • concrete face rockfill • sloping upstream core earth and

rockfill

5.3 Liquefaction of Dam Embankments and Foundations One of the most critical issues relating i effect of earthquakes on dams is whe liquefaction of the dam or foundation

• central core earth and rockfill

occur, and if so, what the consequences ma|

• earthfill with chimney and horizontal drains • zoned earth-earth rockfill (without

cause of dam failures due to earthquake.

filters) • homogeneous earthfill (without filters). (n) Consideration should be given to the effect of earthquake on the strength of dam foundation, particularly where there are weak, or strain weakening seams of rock or clay in the foundation.

(o) Seed et al (1985) suggest that the slopes of concrete face rockfill dams should be flattened to limit displacements in earthquake. They suggest the following:

Historically, liquefaction has been the m|

The following discussion outlines what is m| by liquefaction, how it can be assessed,! measures which can be taken to overcome if

The subject is a broad one, which is subjel continuing research. Some references w|

give overviews include USNRC (1985), Uj

(1989), Finn (1993), Fell et al (1992),J Robertson and Fear (1995 and 1996). 5.3.1 Definitions, and the Mechanics i Liquefaction

The USNRC (1985) gives the folio! definitions of liquefaction and r< phenomena:

36


• "The word liquefaction is used to include all phenomena giving rise to a loss of shearing resistance and to the development of excessive strains as a result of transient or repeated disturbance of saturated cohesionless soils".

(a) Cyclic loading causes densification of dry granular soils by particle rearrangement due to the back and forth straining. If, however, the soil is

saturated and not allowed to drain during cyclic loading, the decreases in

may be considered as the condition when

volume cannot occur and the tendency to decrease volume is counteracted by an increase in pore pressure and decrease in effective stress and

pore pressure equals total vertical stress.

strength.

• "Initial liquefaction" is the condition when effective stress momentarily is zero. This

• "Flow failures" describe the condition where the soil mass deforms continuously under a shear stress equal to the static shear stress

applied to it, eg. slope instability, total bearing capacity failure.

The pore pressures build up gradually with the number of cycles of loading, and only if the pore pressures build up to equal the total stress does the "initial liquefaction" (effective stress <7=0) condition occur.

• "Deformation failures" involve large permanent displacement or settlement, but the earth mass remains stable without great changes of geometry.

Robertson and Fear (1995,1996) point out that flow liquefaction only applies to strain weakening soils, and may occur under static, as

well as cyclic loading. They use the term cyclic softening, split into cyclic liquefaction (in which zero effective stress is reached) and cyclic mobility (in which zero effective stress does not develop) to describe the "deformation failures" condition.

Extensive laboratory testing has been carried out by many researchers to explain the phenomenon of liquefaction. These show that:

(b) The number of cycles (of shearing of the soil) to reach the cf=0 condition depends on the relative density and the magnitude of the cyclic stress compared to initial stress xJa'vo where tc = cyclic shear stress and avo = vertical stress.

Figure 8 shows the results of laboratory testing on a saturated sand which shows these effects. It should be noted that the number of cycles of loading in an earthquake depends on the magnitude of the earthquake. Hence, larger magnitude earthquakes induce more

cycles of shearing, and are more likely to induce liquefactions. This is discussed further in Section 5.3.3.

/


37

Ntimbar of Cyclts, Ne

Figure 8. Cyclic shear stress ratio iJo\a versus number of cycles and relative density to liquefaction from laboratory tests (De Alba etal, 1976; USNRC, 1985),

It will also be apparent that loose sands are more susceptible to liquefaction than dense sands.

(c) Laboratory tests have shown also that the behaviour of saturated cohesionless soils under cyclic loading is dependent on stress history, initial stress

conditions (eg. isotropic or anisotropic loading), and particle grading and assemblage. It is difficult to simulate field behaviour in the laboratory, mainly because it is very difficult to obtain undisturbed samples of the soils for testing.

Soils which have been subject to liquefaction do retain some strength. An understanding of the

stress-strain relationship of these soils in undrained loading is critical to the understanding of liquefaction, and is described

As the soil is sheared, positive pore prs are generated as the soil tries to conti volume, but cannot do so in the und condition. In the monotonic loading cai

available strength is greater than the statii stress, so if this was representative of a

the slope would be stable. On the right: Figure 9(a) is shown the effect of cyclic li from an earthquake. If this induces sui strain, the available undrained strength s less than the static shear stress, anc

continuing deformation will occur und static loading, with further reduction in st until the residual undrained strength is r< (Sus). (Sus is also known as the stead; undrained strength or "residual strength") is a case of flow failure, and would likel) in large deformations in a dam.

(1985). Figure 9 illustrates the important

Figure 9(b) illustrates the stress strain beh of a saturated cohesionless soil loaded undrained state, where the soil is dilaf

points.

behaviour during shearing. In this case,

in some detail in USER (1989(a)) and USNRC

monotonic undrained loading, negativf Figure 9(a) illustrates the stress-strain behaviour of a saturated cohesionless soil loaded in an undrained state, where the soil is contractive in

behaviour and strain weakening (at large strains) during shearing. The plots assume that

pressures are generated, so the soil contir

increase in strength until a maximi reached. In this case, cyclic loading le larger strains, but a stable, strain har

loading condition remains at the end of cy

the loading begins at the static shear stress Td.

Figure 10 shows conditions where such a static shear stress may occur.

38


Cyclic Loading

(a)

INSTABILITY and FLOW

i

1

(b) DEFORMATIONS of STABLE SOIL T = Shear Stress f = Shear Strain Td = Static (Driving) Shear Stress Sus = Undrained Steady State Strength Figure 9. (a) Unstable; and (b) Stable behaviour under static and cyclic loading (Castro, 1976;

USNRC, 1985). Whether or not a soil will be contractive or dilative in behaviour depends on its void ratio (which is related to the relative density and the stress conditions). Whether it will reach a flow failure condition will depend on the static shear stress, the undrained strength vs strain behaviour, and the amount of strain induced by cycling of load. In practice, some drainage may occur during

the earthquake loading, particularly if the soils are highly permeable, and/or if the drainage path is very short. This drainage dissipates pore pressures, reduces strains, and hence lessens the

likelihood of a flow failure condition.

It should be noted that: • Soils which have liquefied, can liquefy again, so one cannot rule out the possibility of liquefaction on this basis. • Liquefaction is an issue in Australia - it has been observed during earthquakes in the South Australia-Victoria border region, and the larger magnitude Australian earthquakes are quite sufficient to cause liquefaction. • Australian earthquakes may have somewhat

different characteristics to those in other countries, but it is considered that the methods for assessing liquefaction potential described below are sufficiently accurate for use in Australia.

/


39

(a)

FAILURE SURFACE

SAND

(b)

IMPERMEABLE SOIL lc) . . . SAND'•: . .

Figure 10. Examples of situations involving the existence of static shear stress in the soil: (a): ground; (b) embankment on level ground; (c) earth dam. 5.3.2 Soils Susceptible to Liquefaction Saturated sands, silty sands, silts and gravelly sands are susceptible to liquefaction. Figures 11 and 12, reproduced from USNRC (1985), show the particle size envelopes for potentially liquefiable natural soils, and mine tailings. However, later experience has shown that even

soils with small amounts of clay may liquefy.

40

It can be seen that the mine tailings ar susceptible to liquefaction than natura possibly reflecting their uniform size am deposition. Troncoso (1990) and Tronco (1988) present some evidence that tailir "age" and develop greater resistai

liquefaction with time. The behavi tailings can be considered to be represe of dredged fills.


Figure 11. Limits in the gradation curves separating most liquefiable and potentially liquefiable soils. (Tsuchida, 1970; USNRC, 1985). Note that soils outside these boundaries may liquefy.

Generally the presence of fines (silt and clay size particles passing 0.075mm sieve) reduces the susceptibility to liquefaction. USSR

clay content as % by weight passing 0.005mm, not the usual passing 0.002mm

(1989(a)) indicate that soils with clay fines (as

definition)

opposed to silt) are susceptible to liquefaction, but the soil will not be subject to liquefaction if:

They indicate that soils are potentially

liquefiable if Clay content >20% or Water content <0.9 (liquid limit)

(Note USSR (1989(a)) define

<15% <35% Water content >0.9 (liquid limit) Clay content

and or

Liquid limit

Figure 12. Ranges of grain sizes for mine tailings slimes with low resistance to liquefaction. (Ishihara,

1985, USNRC, 1985).


41

5.3.3 Seed Method for Assessing

Liquefaction

where tav = average peak shear stress a^ = maximum acceleration ground surface

The most widely accepted, simplest and most practical method of assessing whether there is a

g'0 = effective overburden stre;

depth under consideration

potential for liquefaction for horizontal ground conditions, has been developed by Seed and his co-researchers over many years. The method is semi-empirical, and is based on the maximum acceleration induced by the earthquake a,,,.,^ the SPT 'N' value corrected for the SPT hammer

ct0 = total overburden stress al

same depth rd = stress reduction factor = ground surface and 0.9 at 1

energy and for overburden pressure (Ni)60,

earthquake magnitude (M), and fines content of the soil (% passing 0.075mm). It is based on recorded cases of liquefaction during earthquakes in USA, Japan and China. Details are given in Seed and De Alba (1986) and in USNRC (1985). The method is recommended for the initial assessment of whether a soil will liquefy.

g = acceleration due to grs

(9.81m/sec?) (c) Determine the stress ratio (t, which will lead to liquefaction, by following procedure: (i) Determine (N,)^, from the measi SPT 'N' value, corrected to 60% em

ratio from Table 12 and to 100

The steps are:

overburden stress using Figure 13. (a) Estimate a^ for the site, ie. maximum acceleration at the ground surface

N, =CnN,

ie.

during the design earthquake.

ERm

and

N,60

(b) Estimate average cyclic stress ratio Tav/a'0 induced by the earthquake from

60

where Nm = measured SPT valu

0-65amaxcTorrf

=

:N„

ER,,, = measured rod enerj

Vog

Table 12. Summary of Energy Ratios for SPT Procedures (Seed and De Alba, 1986). Country

Hammer

Hammer Release

Type Japan

Donut Donut

Estimated Rod Correction Factor fl

Energy (%) Free-fall

Rope and pulley with

60% Rod Energy f

78 67

78/60= 1.30 | 67/60 = 1.12 J

60 45 45 60 50

60/60= 1.00 I 45/60 = 0.75 J

special throw release

United States Argentina

China

Safety Donut Donut

Rope and pulley Rope and pulley Rope and pulley

Donut Donut

Free-fall

Rope and pulley

(a) Japanese SPT results have additional correction for borehole diameter and frequency effects.

42

45/60 = 0.75 1 60/60= 1.00 1 50/60 = 0.83 1

(b) The safety hammer is the preva| method in the United States.


(c) Pilcon-type hammers develop an energy ratio of about 60%. Australian hammers are generally activated by a trip deyice and are free

In the absence of measurements, it will

probably be reasonable to assume that Australian free fall hammers have an energy ratio of 60%. It is strongly recommended that the actual energy ratio

fall. Some recent measurements in

be measured where liquefaction is being

USA reported in Drumrightet al (1996)

assessed.

indicates free fall US safety hammers gave an energy ratio of 0.9 to 0.95, and

reported tests by Kovacs (1994) showing energy ratios of 0.8. Tests on hammers used recently on an

(ii) Determine which will lead to liquefaction for a magnitude 7.5 earthquake (from Figures 14 or 15.

Australian Dam gave Er « 59% to 61%.

Figure 14 applies to sand with less than 5% fines passing 0.075mm. Figure 15 applies

Robertson and Fear (1996) recommend that Er be measured directly because it is influenced by the type of equipment and its condition.

where there are more fines. Note that the fines

reduce susceptibility to liquefaction. The influence of clay in the fines is discussed in Section 5.3.2.

Cm

Figure 13. Graphs for value of overburden correction factor CN (Seed and De Alba, 1986).

/


43

OjEt

OS

v0"

0.3

.02

FINES CONTENTS 5* Omm C«* Id* cMM>0) • jjquiiocScn > Iw^BeNan Pm+minca\ Mo • « Jogonttt daw • ¦ ( CltMH doll , A ^ 4

10

20 30

«*0

40

»

Figure 14. Relationship between stress ratios causing liquefaction and (N,),# values for clean sa magnitude 7.5 earthquakes. (Seed and De Alba, 1986).

20 30

Figure 15. Relationship between stress ratios causing liquefaction and (N,)^ values for silty sa magnitude 7.5 earthquakes (Seed and De Alba, 1986). (iii) For earthquakes of magnitude other than 7.5, correct the values of Tav/a'0 by

the factors in Table 13.

44


Table 13. Representative Number of Cycles and Corresponding Correction Factors (Seed and De

Alba, 1986). Earthquake

Magnitude (M)

Number of Representative Cycles at 0.65 t^

Factor to Correct Abscissa of Curve in

6.0

5-6

Figures 14 and 15 0.89 1.00 1.13 1.32

5.25

2-3

1.5

8.5 7.5

6.75

26 15 10

It should be noted that the magnitude of the earthquake has a significant influence on whether liquefaction will

Salmon (1995) points out that when earthquake loading is determined by a probabilistic analysis of the history of

occur. It may be assumed that

earthquakes in the seismotectonic zone

liquefaction will not occur for earthquakes of Magnitude 5 or less

around the dam (as is done in Australia), it is possible to determine the contribution of different magnitude

(there are not sufficient cycles of

loading for small earthquakes). However, where static liquefaction may occur, even small earthquakes could

trigger failure. Morgenstem (1995) discusses static liquefaction which is likely to occur only in very

earthquakes to the assessed ground motion. Figure 16 gives an example

which shows that the major contribution to the estimated ground motion for a given PGA (in this case

the 1 in 1000 PGA) is from small

loose/poorly compacted, saturated soils

magnitude earthquakes near the dam.

including dredged sand, mine

Given the nonlinear relationship implied in Table 13, and Figures 14 and 15, this has an important influence on the assessment of the probability of liquefaction, and should, if practicable,

overburden dumps and mine tailings.

be included in the assessment.

/


45

50

MAGNITUDE CONTRIBUIIONS

GO OS

I » t

Sums5 6

o

10

lb.

-t-

3 4 5 6 7

MACNIIUOE (M) 50

oisiANcr coNfRinunnN^ Nnle ConUfbolKjftS /ro
Suniat9 0

TVvn^. 0 20 40 60 SO 100 120 140 160 ISO 200 220 240

OISIANCC (km) Magnitude-distance contributions (Murrin Substation) From shallow seismogemc zones lo PGA at P-0 001

Figure 16. Magnitude-distance contribution (Sail (d) Compare the estimated ija'0 for the design earthquake with that required to cause liquefaction.

The "factor of safety" to be applied to this depends on the degree of conservatism in selecting a^ and may

be between 1 and 1.3 (on tjo'0).

i,1995). experience on two dam sites in Australia s the SPT 'N' value vs qc relationships shoi Figure 17, and implied in Figure 18 d< apply, so considerable care should be tak using CPT values. A further problem is th; unlike the SPT, where a sample is recover! (and a % fines can be determined direct the SPT (N,)^ value) for CPT, fines co must be determined from other data,

The Seed approach has been extended to use of the Static Cone Penetration test (CPT) by correcting CPT q,. values to SPT (N,)60 using Figure 17. Alternatively Figure 18 can be used

brings in a potential for error in the assess The CPT does have the advantage of pro continuous data, so a good approach is '

both CPT and SPT on the site.

in lieu of Figure 15. However, recent

46


- 4 o a. c o

2 3-

O JomfoUowtki tf oL 09091 0 MurorrocM and KoboroiM UM2) A (ihihofo and K090 (Oil) it Robvrlion (1902) V MHehtll (1903) O HoriNr •> oL (1964)

0

ne •»' bkul/lool

OOI 002 003 01 0 2 0 5 I Mean Groin Sit*, Oj0 -mm

Figure 17. Variation of ratio with mean grain size (qc) measured in tsf «100 kPa) (Seed and De

Alba, 1986).

-i 1 1—iM> 7.3 •orthquaku

06

% riots >33 213 slO 13 Ojolmm) 01 0.2 0.2302304 08

V 0'

le/Ngo 3.3 41 4.4 4,4 4^33

Z 03 >» u o^

o.i

_1_

X

40 80 120 160 200

240

Modifiad Con* P*n*tralion R*«iilanc*. qe|-l«f

Figure 18. Relationship between stress ratio causing liquefaction and cone tip resistance for sands and

silty sands (1 tsf»100 kPa (Seed and De Alba, 1986). There are several problems in applying the Seed et al semi empirical approach:

applications it can therefore only be directly applied to assess whether liquefaction would occur without the

(a) It has been developed for level or near

dam.

level ground conditions. For most dam


47

Seed and Harder (1990) recommend that to allow for the static driving stress

initial effective stresses, soils will be dilative or more contractive, and

values will apply.

due to the dam, a correction factor

should be applied to the calculated

(xayCT'0)a^=(xav/a'0)a^ • Ka

It will be noted that for soils y relative density greater than 45%, greater than unity, so the effect (

where Ka is a correction factor

likelihood of liquefaction.

(Tav/CT'0). This is calculated using

in-situ shear stress is to lessei

determined from Figure 19. To determine Ka, the relative density and a (the ratio of static driving shear stress on a horizontal plane to the initial

(b) Seed and Harder (1990) recommend that for effi overburden stresses greater thar kPa, a further correction is requii

effective overburden stress, (tav/a'0) has element analyses.

the calculated xja'^ Thisiscalci using

The correction only applies to soils

(tav/o' 0)0*0 = tfo o)&o= lOOkPa * KJ

to be determined, eg. from finite

where Tav/a'0 <300 kPa. At higher where K^ is a correction factor determinec a',, and Figure 20.

2.0

Ora. 55 - 70%y

X

Thv

oc

Thv

Op ^ BOOkPa 0.1 0.2 0.3 0.4

0.5

OC Figure 19. Correction factor K,, to account for static driving shear stresses (Seed and Harder, 1990

48 ANCOLD Guidelines for Design of Dams for Earthquake

i

values are variable (over that increase

of potential liquefaction intervals indicated is great enough to be of

which occurs with the increasing

concern.

(c) In most natural soil deposits the SPT

overburden stresses). There are no

clear guidelines given by Seed et al as to how this should be accounted for.

• If the deposit being sampled is

USSR (1989(a)) discuss this issue and present some useful practical points which are recommended for use. These

increase the blow count, implying a more dense, less potentially liquefiable soil. To check for this,

include:

USBR (1989(a)) recommend

known to contain, or may contain gravel, these coarse particles may

recording SPT blow counts for each 300mm of penetration, and correcting for the erratic effects of

• The results of each interval of each drill hole, with regard to liquefaction potential, should be prepared in a table and should be presented on geologic cross sections and profiles to allow examination of the frequency and continuity of those intervals indicating liquefaction susceptibility. From such a presentation a judgement is drawn as to whether or not the continuity

gravel (as a minimum, the three sets

of 150mm blow counts should be checked to see whether this potential irregularity is present). Where gravel is extensive, shear wave

velocity methods should be used to assess liquefaction potential.

X

}* *x

V

X X

X X

X

•5s .* *

<

t X

;X

X : < X

X

:[

X

0 100 200 300 400 S00 600 700 800 EFFECTIVE CONFINING PRESSURE (To' kPa Figure 20. Correction factor K, to account for in-situ effective stress ct'0 greater than lOOkPa (Seed and

Harder, 1990). Fear and McRoberts (1995) reconsidered the the Seed method used average SPT 'N' values in Seed database, and determined that in general, the critical zones. The USBR procedure


49

discussed above effectively does the same

figure 21, t,. represents the limiting shear strain

thing, but this emphasises that it would be over

at liquefaction. The 3% line is equivalent to the design curve in Figure 14. It will be noted there are changes in the graph at the low SPT values, Robertson and Fear (1996) suggest that the fines content can be allowed for by adding the following corrections to (N,)^.

conservative to use the minimum 'N' value in

the Seed method. Fear and McRoberts (1995) present revised graphs of liquefaction potential based on the use of minimum 'N' values. These

would need to be used with caution since it is not uncommon for individual SPT tests to be affected by loosening in the drilling process giving unrealistically low values.

ACNOeo = 7 ifFC > 35% =0 ifFC <5%

Robertson and Fear (1996) indicate that at the NCEER Workshop it was agreed that the curves shown in Figure 21 should be adopted in preference to those in Figures 14 and 15. In

ACNOao = (FC-5) if5%
ctOr :%>Ejlimettd rqn^t (S64>

in I =

PropoiH

for

Band en ttit dote By

by Uti 119791

ToiumcViu end Yoihin

(13041

_ Ol — i"iqn esireî coKnnei —j-iii»m«3cl«T"No ti^mlieam dome?!--

0.6 Licui'eetien witn

7^-20% -10% -a* 0 5-

Ne

C-f-

3k <=1

o.£-

0 2-

01

10

20

30

40

50

Figure 21. Recommended cyclic resistance ratio (CRR) for clean sands under level ground conditions based on SPT (Robertson and Fear, 1996).

50


It is important to note that a positive result from

the Seed method only indicates that a soil may be susceptible to liquefaction. It does NOT mean that just because (part of) the dam foundation will liquefy, the dam will fail. What should follow such an outcome, is an assessment of the extent and continuity of

In practice, this will leave many sites in the "grey" zone.

Shear wave velocity may also be used to assess

liquefaction by relating peak ground

potentially liquefiable soil, and the residual

acceleration, and a history of performance of sites in earthquakes. Bierschwale and Stokoe

undrained strength of that soil, for use in post liquefaction analysis.

(1984) and USNRC (1985) give details. USNRC (1985) also gives details of a method

Liao, Venziano and Whitman (1998) provide some useful information which will help in the application of liquefaction in a probabilistic

which assesses peak strains due to the earthquake from the shear wave velocity, and

compares this to threshold strain. Brief details of these methods are given in Fell et al (1992).

framework.

5.3.4 Shear Wave Velocity Methods for Assessing Liquefaction Potential

5.3.5 Determination of Residual Undrained Strength In recent years there has been quite a lot of

Shear wave velocity is affected by many of the variables which influence liquefaction, eg. (relative) density, confining pressures, stress history, geologic age. Hence, it has some use as

an indicator of potential for liquefaction.

discussion of the post liquefaction condition. This is usually discussed in terms of "residual (undrained) strength", "field residual strength", or "steady state undrained strength". Some of these are expressed as plots of residual undrained strength versus SPT 'N' value (eg.

The shear wave velocity may be obtained by

Seed (1987), Lo and Klohn (1990), Seed and

downhole, crosshole or surface to downhole seismic methods, or by a seismic cone

Harder (1990). These plots are developed from backanalyses of liquefaction failures. Figure 22 is the plot presented by Finn (1993).

penetration test (a modification of the piezocone test) developed by Campanella et al

(1986).

Finn (1993) indicates that lower bound strengths are often used for these analyses,

As discussed above, some coarse grained

although 33rd percentile values have sometimes

cohesionless materials (gravels, cobbles) suspected of being potentially liquefiable cannot be successfully sampled using SPT. If

percentile gives very low (-zero) residual undrained strengths at SPT less than about N =

crosshole shear wave velocity data have been

6.

obtained on the materials, and these data accurately represent the deposits in plan and

been used. In either case, the lower bound or 33

making a judgement on liquefaction potential.

An alternative approach adopted by a number of authors including Ishihara (1994) and Finn (1993), is to relate the normalised residual

USBR (1989) recommend that:

undrained strength Su/ct'vo (where SU5 = residual undrained strength and ct'Vo effective vertical

section, then they provide a viable means for

• if shear wave velocities are >365m/s, the

deposits may be judged non liquefiable if shear wave velocities are between 245 and 365m/s, the deposit may be considered likely to be non liquefiable, but supporting evidence should be obtained

stress) to either SPT (N,)^ values (N values corrected to 100 kPa effective stress, 60% energy ratio hammer) or to cone penetration resistance qc.

• if shear wave velocities are < 245m/s, the

deposit may be judged liquefiable.


51

—

2000

—<

t»»TM0U»«t " INDUCCO ^PJ"!»V"0»tVt0|>?
I60C H O Z tu cr

O

CANTHOUARC - IMOUCO LKiucrwriM ultT' JPT 3ATJL AMD KtJIOUAI. MKA-tTim «« •"* WTI»««I. CO NTT RUCTION - INOUCCO UIOUtfieTION **8 tUUMt MJt KUTOmtl.

1200

e < u a 800 UJ z < s Q Z 3

400 io»m un r*»>»»Noo OA"

-I

< O

a 35 4 8 12 16 zo EQUIVALENT CLEAN SANO SPT BLOWCOUNT,

Figure 22. Correlation of residual undrained shear strength with SPT (NOso (Finn, 1993). Finn (1993) indicates that values of Sus/a'vo around 0.06 to 0.1 have been used for the design of a number of dams. Ishihara (1994) presents data from backanalysisof field failures showing

Stark and Mesri (1993) argue that these strengths obtained from case histories are sometimes affected by drainage, and hence should not be used where drainage may not

a lower limit of Su/a'vo of 0.05. He indicates

occur. They recommend use of a formula

that laboratory tests, particularly on reconstructed samples, gives lower bound

relating SJ
results.

number of researchers. Their recommended

equation is:

Fear (1996) reviewed the field data used by SuMo = 0.0055 (NJXsoc

Seed and Stark and Mesri (1993). This shows that the data used to produce Figure 22 is very approximate. Many SPT 'N' values were only

where (N,)'eocs = SPT 'N' value corrected to

estimated, and a variety of analysis methods with often poor modelling of geotechnical conditions were used to determine the residual undrained strength. The range shown in Figure 22 is a true reflection of the data.

100 kPa overburden stress, and 60% energy ratio, for clean sand (cs). For soils with fines (passing 0.075mm), they recommend correcting the measured (NOeo values by adding A^,)^ taken from Table 14. This correctipn is similar to that detailed above.

'-vYtV/'V "

52

ANCOLD Guidelines for Design of

uake

Table 14. Correction to SPT (Ni)60 Values to Allow for Fines (Stark and Mesri, 1993). Fines Content

10 15 20 25 30 35 50 75 There is a diversity of views on what strength to use, particularly at low SPT (N,)^ values. All that can be recommended is that those involved read the relevant papers and make some

judgement for the site under consideration. Of the available methods, that proposed by Stark and Mesri (1993) is the most complete, and is recommended. For silty sands and sand with some silt, or interlayered sand with sandy clay,

silt or clay, drainage during earthquake is unlikely to occur, so the Stark and Mesri method is likely to give reasonable estimates. For uniform clean sands which may drain it is

2.5

4 5 6 6.5

7

7 7 (a) Changing operational procedures for the project, eg:

(i) lowering the maximum water level in the reservoir

(ii) limiting public access to the area downstream of the dam

(iii) institute early warning systems downstream.

(b) Improve in-situ foundation conditions to

reduce liquefaction susceptibility including:

possibly on the conservative side. It is reasonable to expect that there will be more

(i) excavation of existing materials of potentially liquefiable soils of inadequate relative density sand, and

papers published on this topic, as it is an

replacement with compacted non

important one receiving the attention of a number of prominent researchers. Some

liquefiable soils, eg. replacement of silt with well compacted sandy gravel,

information in Fear (1996) seems to indicate that at low SPT 'N' values, Stark and Mesri is generally acceptable, but might not be

sand, or gravel

(ii) densification and increase in in-situ

conservative in particular situations, which

lateral stress by in-situ methods:

cannot be defined other than by laboratory testing. At high SPT 'N' values, Stark and Mesri's (1993) method is likely to be

blasting, compaction piles; vibroflotation or vibratory probes; compaction grouting; dynamic compaction (consolidation). These methods are more likely to be applied

conservative, and laboratory testing is needed to determine the actual strength.

The use of analysis for residual undrained strength is discussed further in Section 6.4. 5.3.6 Measures to Improve Liquefaction Resistance

USNRC (1985) reproduced in Fell et al (1992) discuss methods for altering the hazard, reducing the risk of liquefaction or improvement of stability during and after earthquake. These may apply to an existing dam or to new dam construction. USNRC

(1985) suggest there are four general classes of mitigation measures:

to new construction than to remedial works for existing dams.

Vibroflotation and dynamic consolidation are relatively common procedures, and particularly in the case

of vibroflotation, high relative densities

(70% to 85%) can be achieved at moderate cost to depths of up to 30m in clean sands, sandy gravels and sands

with minor silt. Brown (1977) discusses the applicability of the methods. Figure 23 shows the soils which can be treated by vibroflotation.


53

particle sue - millimetres tilt Uay Mn« |m«dium|coart«

sand |m«dnjH ceont

qravtl r otin fin* |tn«diuin|coars*

Grading of soils which can be d«nsifi«d by vibroflotalion Zon« B — Most suited Zon« C — Can b« treated but finas can eauM probi«ms Zon« A — Can b« (r«at*d bul grav*4 eaus«s w«ar in •quipmtnt and »low progress

Figure 23. Soils suitable for vibroflotation(adapted from Brown, 1977, and industry data). (iii) in-situ improvement by alteration of material, eg. mixing in-place material with additives, eg. lime, cement or

asphalt introduced in augered piles; removing in-place material by jetting and replacing with suitable materials. These methods are costly and unlikely to find wide application (iv) grouting or chemical stabilization. Chemical or cement grouts may be used to improve the strength and stiffness of soils. Generally, grouting of soils is difficult and costly, and not likely to be the economic solution. It also has the disadvantage that permeability is reduced, lessening the ability of the soil to dissipate pore pressures built up by the cyclic loading of an earthquake.

into a new dam design. The' several effects

• the effective vertical stre the foundation (of poter

liquifiable soil) is increa This increases the cyclic strength and shear moduli • the static shear stress©

reduced, improving earthquake stability • the freeboard may increased by raising the level of an existing dam \ berm

(ii) piles, caisons, freezing more likely to be usee buildings than for dams bec| the large costs involved

(iii) rebuilding the structure extreme cases a new dam if

Figure 24 shows schematically some of the

techniques. USNRC (1985) gives additional details of the techniques and the

applicability. (c) Structural solutions: (i) berms - these may be added to an existing dam or be incorporated

54

needed as shown in Figured (d) Drainage solutions - drainage so set out to relieve pore pressuresb

during the cyclic loading earthquake by providing permeable drainage paths, dewatering the soil to a 0 saturated condition


DENSIFICATION AND INCREASE OF LATERAL STRESS - COMPACTION PILES - VIBRATORY PROBE - VIBROFLOTAtlON - COMPACTION GROUTING

MATERIAL IMPROVEMENT - INPLACE MIXING WITH ADDITIVES - INPLACE JETTING AND REPLACEMENT

GROUTING (CEMENTATION) - CHEMICAL GROUTING - PARTICULATE GROUTING

Figure 24. Schematic illustration of the use of some in-situ improvement techniques to reduce the liquefaction hazard for earth dams (USNRC, 1985; Silver, 1985). (i) Pressure relief wells. These may be

of the dam embankment to ensure

conventional water wells with stainless steel well screens, spaced at sufficiently close centres to allow

that so far as practicable the embankment is partially saturated. If

dissipation of pore pressures from the cyclic loading. Since many alluvial soils are stratified, the horizontal permeability is much greater than the vertical, so the drainage wells will significantly improve the drainage. Stone columns may be used in lieu of conventional wells, and may be built

sands, eg. tailings sand are used to construct zones which will become saturated, layers of more permeable

sand or gravel would be included, to dissipate pore pressures.

(iii) Dewatering and air injection. These are shown in principle in Figure 26 and are likely to be used as remedial

by backfilling vibroflot probe holes

measures in existing structures, and

with gravel. However, in this case

they may not be as effective as

then only when other methods are not adequate. The major problem is that pumping or injection of air must be continuous which is undesirable

screened wells.

from dam safety and cost viewpoints.

control of erosion of fines into the stone columns may not be possible and

In addition to helping dissipate pore pressures built up by the cyclic loading, pressure relief wells may be

As discussed above, it may well be good engineering practice to incorporate into the

used to reduce uplift pore pressures under static conditions. Since "liquefaction" is a process of reduction

improve the resistance of a dam foundation to liquefaction, rather than to rely on the ability to predict what will happen in a design earthquake.

of effective stresses to zero by buildup

Positive measures to density loose-medium dense sands, and/or to add a berm to reduce static shear stresses can be carried out at a

of pore pressure, to start at lower static

pore pressures may be critically / important to prevention of liquefaction. (ii) Drainage layers. Drainage layers, ie. horizontal drains, vertical drains, would be incorporated into the zones

design one or more of these measures to

relatively small incremental cost to a project, and improve the degree of confidence of the safety of the dam very significantly. In particular the risk of flow failure can be considerably reduced.


55

/.////////////m/MW/M/SMmmrm ADtOUATf FOUNDATION MIMIC STAilLfTY

AOCOUAT* FOUNDATION ttltMIC JTA1IUTY

INCREASE FREEBOARD WITH BERM

CONSTRUCT DOWNSTREAM EARTH BERM

DOWNSTREAM BOLSTER SECTION - ROLLCRETE STRUCTURE THROUGH THE DOWNSTREAM TOE - REINFORCED EARTH STRUCTURE THROUGH THE DOWNSTREAM TOE

NEW DOWNSTREAM CROSS SECTION - PHYSICAL SEPARATION FROM EXISTING CROSS SECTION

Figure 25. Schematic illustration of of the use of structural measures to reduce the liquefaction hazard for earth dams (USNRC, 1985; Silver, 1985). Ledbetter(1985) also gives some details on methods of reducing liquefaction potential of foundations.

AIR HfAOCR HOKIZONTAl ORAMACI

!LiL

T&77777?'/,, > / ///////////////7////// // ////,

'////////W,/////, V tTOtd STONE COLUMNS

INJECTION OF AIR — PARTIALLY SATURATE LIQUEFIABLE MATERIALS

STONE COLUMNS

KAVINOHCAOCflS

•KINC tint

f ARTIALLY tATURATfD (OILS

DEWATERING - PARTIALLY SATURATE LIQUEFIABLE SOIL

FREEZING - FREEZE SOILS THAT CAN LOOSE STRENGTH

Figure 26. Schematic illustration of the use of groundwater control measures to reduce the liquefaction hazard for earth dams (USNRC, 1985; Silver, 1985).

56


6. SEISMIC STABILITY

• The most susceptible to failure under

ANALYSIS OF EMBANKMENTS 6.1 Preamble

because these are susceptible to liquefaction if the fill is granular and saturated. • In dams constructed of saturated cohesionless soils, the primary cause of

The methods of analysis currently used in practice to evaluate seismic stability of

• There are very few cases of dam failures

earthquake loading are hydraulic fill dams,

damage or failure is the buildup of pore water pressure (liquefaction) under the earthquake loading.

embankment dams vary widely, ranging from

during the earthquake shaking. Most of the

simple limit equilibrium type analyses to highly sophisticated numerical modelling techniques. For the purposes of these guidelines they are classified as:

failures occur from a few minutes up to twenty-four hours after the earthquake. However, cracking and displacements do occur during the earthquake.

• Pseudo-Static Analysis

Based on the above and similar observations,

• Simplified Methods of Deformation Analysis • Post Liquefaction Analysis • Numerical Modelling Techniques

the US Bureau of Reclamation (1989) classifies embankment dams into two main categories: (1) not susceptible to liquefaction and (2) susceptible to liquefaction For the purposes of seismic stability assessment, they also identify two types of analyses: deformation analysis and post liquefaction (earthquake) analysis. They recommend that deformation analysis be performed on dams not susceptible to liquefaction and post earthquake stability analysis be performed on dams susceptible to liquefaction. This guideline recommends that a similar approach be adopted but adds the proviso, that where significant strain weakening may occur due to displacements induced by the earthquake, post earthquake stability should be

- Total Stress - Effective Stress

In this section, firstly, a brief description of each method of analysis is provided, and then a set of guidelines is outlined for seismic stability assessment of embankment dams in Australia.

The simplified methods of analyses including pseudo static, and post liquefaction, rely heavily on the lessons learnt from the performance of dams during past earthquakes. Major reviews of past performance have been conduced by

Sherard (1967), Sherard et al (1974), Seed (1979, 1983) and Seed et al (1978, 1985). These studies show that:

done taking account of the strain weakening.

6.2 Pseudo-Static Analysis

• Even at short distances from the epicentres, there have been no complete failures of

Up until the 1970s, the pseudo-static analysis was the standard method of stability assessment

embankments built of clay soils, but several

for embankment dams under earthquake

dams have come close to failure.

• Well constructed dams of clay soils on clay or rock foundation not susceptible to strain weakening, can withstand extremely strong

shaking resulting from earthquakes of up to magnitude 8.25 with peak ground accelerations ranging from 0.35g to 0.8g.

(They suffer cracking, but few, if any, have failed catastrophically). • Dams which have suffered complete failure as a result of earthquake shaking have been / constructed primarily with saturated sandy materials or on saturated sand foundations.

Liquefaction is a major contributory factor in these failures.

loading. The approach involved a conventional limit equilibrium stability analysis, incorporating a horizontal inertia force to represent the effects of earthquake loading. The inertia force was often expressed as a product of

a seismic coefficient k and the weight of the sliding mass W as shown in Figure 27. The larger the inertia force, the smaller the safety factor under the seismic conditions. In this approach, a factor of safety (FOS) of less than one implied failure, whereas F0S>1 represents seismically safe conditions, although as shown in Table 15, a higher factor of safety was often utilised to take into account material degradation under cyclic loading.


57

Figure 27. Pseudo-static method of assessing seismic stability of embankments. The seismic coefficients used in this approach were typically less than 0.2 and were related to the relative seismic activity of the areas to which they apply. In the United States, for

example, they ranged from 0.05 to 0.

Japan they have characteristicallybeen 1< about 0.2, and similar values have been

other highly seismic regions through world, as shown in Table 15.

Table 15. Seismic coefficients used in selected embankment dams (Seed, 1979).

Dam

Country

Horizontal seismic

coefficient Aviemore Bersemisnoi

New Zealand Canada

Digma Globocica

Chile

Karamauri

Kisenyama Mica Misakubo Netzahualcoyote Oroville Paloma Ramganga Tercan Yeso

Yugoslavia Turkey Japan Canada Japan Mexico

USA Chile India Turkey

Chile

0.1 0.1 0.1 0.1 0.1

0.12 0.1

0.12 0.15 0.1 0.12 to 0.2

0.12 0.15 0.12

For dams not susceptible to liquefaction, Seed (1979) suggested a seismic coefficient of 0.1 g for magnitude 6.5 earthquakes and 0.15g for magnitude 8.25 earthquakes to obtain a safety factor of 1.15. However, he qualified this recommendation to point out that it applied to

The US Army Corps of Engineers (198' extended the basic pseudo-static method^

"most" earthquakes, and was "often adequate".

conditions for cohesive soils and < conditions for free draining granular mi with a 20 percent strength reduction to al strain weakening during the earthquake 1< They require a factor of safety greater th If a dam fails to satisfy this, more accur

The cases shown in Table 15 are all relatively old, and of interest only in giving the history of developmentof the methods of analysis.

58

as a screening method for dams not susc to liquefaction. They recommend us seismic coefficient equal to one-half c ground acceleration and the use of uni


detailed analyses are required. Their approach has been calibrated against a large number of deformation analyses, and they state that up to 1m of deformation may occur.

strength. This near-disastrous event, more than any other single event, resulted in an extensive re-appraisal and gradual demise of the pseudo-static analysis.

The pseudo-static method of analysis, despite

Today, it is generally accepted that the

its earlier popularity, was based on a number of restrictive assumptions. For instance, it

pseudo-static analysis is not an accurate tool of seismic stability assessment of embankment dams, and that it may only be used as a

assumed that the seismic coefficient acting on the potential unstable mass is permanent and in one direction only. In reality, earthquake accelerations are cyclic, with direction reversals. Therefore, the concept it conveyed of earthquake effects on embankments was very inaccurate. Also, the concept of failure used in

screening tool (for dams not susceptible to liquefaction). It is recommended that the US Corps of Engineers (1984) method may be used as a screening method for well constructed earth

and earth and rockfill dams, which are not susceptible to liquefaction or significant strain

the approach was influenced by that used in static problems. It is clear that a factor of safety

weakening in the dam or its foundation.

of less than one cannot be permitted under static

6.3 Simplified Methods of

conditions as the stresses producing this stage will exist until large deformations change the

Deformation Analysis

geometry of the structure. However, under

6.3.1 Initial Screening

seismic conditions, it may be possible to allow the FOS to drop below one, as this state exists

The US Bureau of Reclamation (1989)

only for a short time. During this time, earthquake induced inertia forces cause the potential unstable masses to move down the slope. However, before any significant movement takes place, the direction of the inertia forces is reversed and the movement of the soil masses stop and once again, the FOS rises above one. In fact, experience shows that

a slope may remain stable despite having a calculated FOS less than one and it may fail at F0S>1, depending on the dynamic

recommend that for dams not susceptible to

liquefaction, dynamic deformations should not be a problem, and need not be analysed for such

if the following conditions are satisfied. 1. The dam is a well built dam (densely compacted), and peak accelerations at

the base of the dam are 0.2g (gravity) or less; or the dam is constructed of clay soils, is on clay or rock foundations, and peak accelerations are 0.3 5g or less.

characteristics of the slope-forming material.

2. The slopes of the dam are 3:1 (H:V) or

The deficiencies associated with the

3. The static factors of safety of the critical failure surfaces involving loss

flatter. pseudo-static analysis were clearly

demonstrated during the San Fernando Earthquake (M=6.6) in 1971. In this earthquake, the Lower San Fernando Dam

experienced a massive slide in its upstream shell. This slide was very significant as the seismic stability of the Lower San Fernando Dam had been evaluated only five years before the earthquake and a number of reputable design agencies had concluded that the dam was safe against any earthquake that it might be subjected to. Clearly this was not the case and the dam failed despite having a pseudo-static FOS of around 1.3 because liquefaction occurred, with resultant large loss of shear

of crest elevation, (ie. other than the

infinite slope case) are greater than 1.5 under loading conditions expected prior to an earthquake.

4. The freeboard at the time of the earthquake is a minimum of 2 to 3 percent of the embankment height (not less than three feet (0.9m)). Fault displacement and reservoir seiches with regard to freeboard should be considered as separate problems. These criteria are quite conservative,

particularly for dams with rock fill zones, and


59

are seldom met. However they, along with the

potential deformations of an embankme

US Corps of Engineers (1984) pseudo-static method, may be used to determine whether

earthquake loading (Newmark, 1965). contribution, sliding of a soil mass along failure surface was likened to slippitf

more detailed analyses are required. If they are, the normal second step for dams not susceptible

to liquefaction will be to use the Newmark

block on an inclined plane. He envisa failure would initiate and movements^

(1965) and Makdisi and Seed (1978)

begin when the inertia forces exceed V

approaches. Care needs to be taken with screening methods if there are strain weakening

resistance, and that movements woul when the inertia forces were reversed. T

soils in the dam or its foundation.

proposed that once the yield accelerate the acceleration time history of a slippirf are determined the permanent displac can be calculated by double integratf acceleration history above the

6,3.2 The Newmark Approach In 1965, Newmark introduced the basic elements of a procedure for evaluatingthe

acceleration, Figure 28.

Disp. of Soil Block' Toward

Bluff

Figure 28. Double integration method for determination of the permanent deformation period embankment, (Newmark 1965). According to this approach, the permanent deformation in a sliding mass is a function of:

The duration of the earthquake, whic function of the magnitude of the earthc Yield acceleration of the potential s

• The amplitude of the average acceleration time history of the sliding mass, which in turn is a function of the base motion, the amplifying factor of the embankment and the location of the sliding mass within the

mass.

embankment.

60

The validity of the basic principl Newmarks's approach has been demon

by many investigators [eg. Goodman anc

(1966), Ambraseys (1973), Sarma (197 Makdisi and Seed (1978)]. It is generally


29 and the values calculated in steps (a) and (c),

that, provided the yield acceleration is accurately evaluated, the approach can estimate the permanent displacement of a soil mass in reasonably good agreement with those observed during past earthquakes.

(e) Enter Figure 30 with the calculated

However, it should be noted that Newmark's

values of kmax and T0 to determine the horizontal component of earthquake induced permanent displacement, U, in

approach was developed at a time when there

the potential sliding mass.

were no direct methods of computing permanent deformations. Today, deformation

calculations can be made directly using advanced, readily available, numerical codes,

giving a global picture of dam behaviour. It must also be stressed that Newmark's approach

It should be noted that an important step in determination of kn,ax in step (d) is to establish the dynamic properties of the material forming the embankment and the foundation. This can be achieved by:

is limited in application to compacted clayey embankments and dry or dense cohesionless

soils that experience very little reduction in strength due to cyclic loading. The approach should not be applied where embankments or their foundations are susceptible to liquefaction or strain weakening because it will significantly underestimatedisplacements.

• triaxial compression tests, simple shear tests or torsional shear tests conducted under

cyclic loading conditions • resonant column testing • field measurement of shear wave velocities,

either by downhole or crosshole techniques, or

• back calculation using finite element 6.3.3 Makdisi and Seed Analysis

The Makdisi and Seed (1978) approach is based on Newmark's method, but modified to allow for the dynamic response of the embankment as

proposed by Seed and Martin (1966). The approach was developed based on a series of deformation analyses performed on a large

number of embankments subjected to earthquake loading. The approach involves the following main steps: (a) Determine y/h ratio for the potential sliding mass, where y is the depth to the base of the sliding mass and h is the embankment height. (b) Calculate the yield acceleration ky for the potential sliding mass. (c) Determine the maximum crest acceleration U™* and the predominant

period of the embankment T0 (in seconds).

(d) / Determine the maximum value of the acceleration history kmax using the

normalised relationship given in Figure

techniques, modelling measured responses to earthquake events

• empirical relationships, such as those given by Hardin and Drenerich (1972) and Seed et

al(1986). However, given that the laboratory and field based methods are expensive, it is

recommended that the values of kmax be initially calculated using a range of Gmax values obtained

from the empirical relationships. Should the results of the analysis be unacceptable then a more elaborate program of laboratory and/or field testing may be warranted. It should be noted that relating satisfactory dam performance to earthquake induced deformation is very subjective, and generally depends on dam specific criteria about the allowable loss of freeboard, or the tolerable extent of horizontal displacements.

The Makdisi and Seed approach, is widely used and accepted among practising engineers. However, like Newmark's approach, it is

limited in application to dams not susceptible to liquefaction or strain weakening in the embankment or its foundations.


61

v/f

X "N.

Figure 29. Variation of seismic coefficient k,,,,,, with depth of the base of the potential sliding

62


VI -o

c o LJ
3

0.01

0.001

0 0001 0 0.2 OA 0.6 mo*

Figure 30. Variation of yield acceleration with normalised permanent displacement (Makdisi and Seed,

1978). 6.4 Post Liquefaction Stability and Deformation Analysis

dam, and the deformation analysis is performed to obtain a global picture of the liquefaction induced deformation within the dam.

As stated previously, studies of past performance of embankment dams during earthquakes indicate that most cases of dam failures have been associated with dams susceptibleto liquefaction.

The deformation analysis, is increasingly becoming an essential part of a post liquefaction analysis. This is because;, the picture of dam behaviour provided by such an analysis allows the design engineer to make a better judgement

In order to assess post liquefaction stability of embankments by methods such as those described in Section 5.3, two types of analysis are often performed: (1) limit equilibrium analysis, and (2) deformation analysis. The limit equilibrium analysis is performed to calculate post liquefaction factor of safety of the

as to the extent of the remedial measures. A factor of safety alone, in general, is not a very

good discriminating tool for deciding on the extent of remedial work. A factor of safety of 0.9, for instance, can have very different

connotations depending on the geometry and the extent and location of the liquefied zones. It may imply a massive failure or it may simply


63

n result in a limited movement along the slip

Estimates of earthquake induced pore pressi

surface before the embankment attains a new

such as the allowable loss of freeboard and/or

can be obtained through a program laboratory testing and/or the empif procedure proposed by Seed and De (1986). In practice the procedure propos|

the tolerable deformation, provided one keeps

Seed and De Alba (1986) is often preferrej

in mind the limitations of accuracy of the

to its simplicity and ease of use. Cyclic tj of liquefiable soils is very expensive and h| specialised. Furthermore, such tests rdj

stable geometry. But, displacements can be interpreted in terms of performance criteria

methods being used.

6.4,1 Post Liquefaction Limit Equilibrium Analysis The basic steps adopted in a post-liquefaction limit equilibrium analysis are as follows: (1) Identify soils susceptibleto liquefaction within the embankment and the foundation by methods such as those described in Section 5.3.

(2) Evaluate earthquake induced excess pore pressures within the liquefiable materials (note ru = 1 implies liquefaction). (3) Conduct a conventional limit equilibrium stability analysis using the pore pressures calculated in step (2). For liquefied materials use the residual undrained strength as defined and

high quality undisturbed samples, whicj most cases) would be very difficult to obta For many practical problems, the proc|

proposed by See4 and De Alba (1986) degree of accuracy comparable with thjj input data. Seed's procedure involves! following steps: (1) Determine stress ratios t^/ct1,,) the soil layers susceptible! liquefaction (see Section 5.3.3).

(2) Establish N the equivalent numbi effective stress cycles in the df earthquake, (Table 13). (3) Determine the number of cycles! required to cause liquefaction fo| stress ratio calculated in (1).

(4) Using the ratio (N/N,) read! earthquake induced pore pressurej from Figure 31 below.

determined in Section 5.3.5.

0.4 0«

CjcU *0l«, N/N|

Figure 31. Rate of pore water pressure built up in cyclic simple shear tests. It should be noted that in many cases, it will be sufficient to assume that all potentially liquefiable zones (as determined by the methods given in Section 5.3.3) reach the residual undrained strength, with clayey zones retaining full (undrained) strength and rockfill zones

64

retaining full drained strength. In this caS not necessary to evaluate the induced

pressures, and it is sufficient to use the pore pressures at the beginning earthquake.


6.4.2 Post Liquefaction Deformation

Analysis Indicative estimates of potential liquefaction induced deformations can be obtained by performing a static deformation analysis which

used codes within each category is provided in the following sub-sections. 6.5.1 Total Stress Codes

liquefied soils (Finn, 1993 and Khalili, 1994).

The total stress codes, as can be inferred from the classification, are based on the total stress concept and do not take account of pore pressures in the analysis. Therefore, they are

The analysis is often performed in two stages.

used in situations where the seismically induced

incorporates the earthquake induced pore pressures and the residual strength of the

In the first stage, the numerical model is initialised to the pre-earthquake conditions of the dam by simulating the current in-situ stresses. Then, in the second stage, the earthquake induced pore pressures and residual

strengths of the liquefied soils are incorporated into the model to simulate post-liquefaction conditions.

pore pressures are negligible. The total stress

codes may be divided into two main categories: (1) codes based on the equivalent linear (EQL) method of analysis, and (2) fully non-linear codes.

(a) Equivalent Linear Analysis (EQL) The earlier total stress codes were based on the

This type of analysis is also referred to as uncoupled deformation analysis and generally

EQL method of analysis developed by Seed and his colleagues in 1972. EQL is essentially an

leads to conservative estimates of post

elastic analysis, and was developed for approximating non-linear behaviour of soils

liquefaction deformations, as it does not allow for dissipation of earthquake induced pore pressures with time. More accurate estimates of

post liquefaction deformations can be obtained using fully and semi-coupled methods of analysis, as discussed in the following sections.

under cyclic loading. Typical of the EQL codes currently used in practice are: SHAKE (Schnabel et al, 1972), QUAD-4 (Idriss et al,

1973) and FLUSH (Lysmer et al, 1975). SHAKE is a one dimensional wave propagation program, and is used primarily for site response

6.5 Numerical Methods

analysis. QUAD-4 and FLUSH are two dimensional versions of SHAKE, and are used

Numerical modelling techniques such as the finite element method were first applied to the dynamic analysis of embankment dams by Clough and Chopra (1966). This was followed by major improvements by Gaboussi (1967), Schnabel et al (1972), Gaboussi and Wilson

for seismic response analysis of dams and embankments. Given the elastic nature of the EQL analysis, however, these codes cannot take

(1973), Idriss et al (1974), Martin et al (1975), Finn et al (1977), Lee and Finn (1978), White et al (1979), Zienkiewicz and Shiomi (1984), Finn el al (1986), Medina et al (1990) and more recently Li et al (1992). Today, numerical methods are routinely used as both investigative and design tools in many geotechnical

account of material yielding and material degradation under cyclic loading. Therefore, they tend to predict a stronger response than actually occurs. Also, they cannot predict the permanent deformations directly. Indirect estimates of permanent

deformations can however be obtained using the acceleration or stress data obtained from an

earthquake engineering problems.

EQL analysis and the semi-empirical methods proposed by Newmark (1965) and/or Seed et al

The ^dynamic numerical codes used in practice may be divided into two main categories: total

(b) Fully Non Linear Analysis

stress codes, and effective stress codes

(Zienkiewicz et al, 1986 and Finn 1993). A brief discussion of some of the more frequently

(1973).

More accurate and reliable predictions of permanent deformations can be obtained using the elasto-plastic nonlinear codes. Typical of

AN COLD Guidelines for Design of Daws for Earthquake

65

the elasto-plastic non-linear codes used in the

analysis of embankments are DIANA (Kawai,

(utilising multi-yield surfaces) or JM surface theory with a hardening la\3M

1985), ANSYS (Swanson, 1992), FLAG

models are very complex and put ®

(Cundull, 1993), etc. The constitutive models

demand on computing time. JH

used in these codes vary from simple hysteric non-linear models to more complex

especially under strong shakings. Critical

Generally speaking, fully coupled prefM pore pressures under cyclic loadingB complex and difficult. To date, main numerical codes developed in this are® fully validated and still are 'M developmental stages. The validationl performed on a number of these codes I that the quality of response predictil

assessments of non-linear elasto-plastic codes

strongly path dependent (Saada and Bil

can be found in Marcuson et al (1992) and Finn

1987). When the loading paths are sitl the stress paths used in calibrating the tl the predictions are good. As the loadil deviates from the calibration pat! predictions become less reliable. Apal the numerical difficulties, part M unreliability is also due to the poor or la satisfactory characterisation of thi properties required in the models. For iti because of sampling problems, it is ofti difficult to (accurately) determine 1

elastic-kinematic hardening plasticity models. Compared to the EQL codes, the elasto-plastic nonlinear codes are more complex and put heavy demand on computing time. However,

they provide more realistic analyses of embankments under earthquake loading,

(1993). 6.5.2 Effective Stress Codes A major stimulus for the development of the effective stress codes has been the need for modelling pore pressure generation and

dissipation in materials susceptible to liquefaction and thus to obtain better estimates of permanent deformations under seismic loading. The effective stress codes may be

divided into three main categories: fully coupled, semi-coupled and uncoupled.

(a) Fully Coupled Codes In the fully coupled codes, the soil is treated as a two-phase medium, consisting of soil and water phases. Two types of pore pressures are considered, transient and residual. The transient pore pressures are related to recoverable

(elastic) deformations, and the residual pore pressures are related to non-recoverable

(plastic) deformations. A major challenge in fully coupled codes is to predict residual pore pressures. The residual pore pressures, unlike the transient pore pressures, are persistent and cumulative, and thus exert a major influence on

the strength and stiffness of the soil skeleton. The transient pore pressures are cyclic in nature

and their net effect within one loading cycle is often equal to zero. An accurate prediction of residual pore pressures requires an accurate

prediction of plastic volumetric deformations. In the fully coupled codes, this is often achieved by utilising elastic-plastic models based on kinematic hardening theory of plasticity

66

change characteristics of loose sands as n by these models. In general, the accui

pore pressure predictions in fully c models is highly dependent upon the qui the input data. Typical of the fully c codes are DNAFLOW (Prevost,

DYNARD (Moriwaki et al, SWANDYNE (Zienkiewicz, 1991) SUMDES(Lietal, 1992). (b) Semi-Coupled Codes Compared to the fully coupled code semi-coupled codes are more robust art

susceptible to numerical difficulties. Ho they are theoretically less rigorous. In codes empirical relationships such as proposed by Martin et al (1975) and See (1983) are used to relate cyclic strains/stressesto pore pressures. The em nature of the pore pressure generation in

codes generally puts less restrictions on th of plasticity models used in the codes, semi-coupled codes are in general less coi

and computationally demanding. Als( parameters they require are often rou

obtained in the laboratory or in the field. '


is extensive experience in using semi-coupled codes in practice

deformation analysis will be required - if the safety factor is less than one,

Typical of the semi coupled codes are DESRA-2 (Lee and Finn, 1978), DSAGE (Roth, 1985), TARA-3 (Finn et al, 1986), and

FLAG (Cundall, 1993). (c) Uncoupled Codes

perform a Newmark type deformation analysis using the

Makdisi and Seed (1978) approach - should the results of the Makdisi and Seed analysis be unacceptable and further analysis is considered warranted, perform a seismic

In the uncoupled analysis, the pore pressures

are estimated separately using either a program of laboratory testing or an empirical relationship such as the one proposed by Seed et al (1983). Then, they are incorporated into an 3 elasto-plastic nonlinear code to obtain permanent deformations. The uncoupled

analysis is widely used in practice and is generally believed to provide indicative estimates of the post liquefaction behaviour of earth dams. The permanent deformations

obtained using this approach are often on the

response analysis using a fully validated elasto-plastic nonlinear dynamic code.

For dams susceptible to liquefaction perform a post liquefaction analysis consisting of the following main steps: - evaluate earthquake induced pore pressures in the materials

susceptible to liquefaction (note ru = 1 implies liquefaction) - conduct a conventional limiting

conservative side, as the analysis does not allow

equilibrium stability analysis using

for dissipation with time of the estimated pore

the above-estimated pore pressures.

For liquefied materials use the

pressures.

residual undrained strength. For

6.6 Proposed Guidelines

clayey materials consider a strength reduction factor of 15%

The guidelines proposed for seismic stability assessment of embankment dams in Australia

may be best illustrated by the flow chart shown in Figure 32. The procedure involves the following main steps: 1. Establish if the embankment or its foundation is susceptible to liquefaction or not susceptible to liquefaction using the methods detailed in Section 5.3. 2. For dams not susceptible to

liquefaction: - perform an initial screening assessment detailed in 6.3.1

- if the initial screening criteria are not satisfied, conduct a pseudo-static analysis using the US / Army Corps of Engineers Recommendations

- if the safety factor is greater than one, the earthquake performance of

the dam is acceptable and no


67

Figure 32. Embankment seismic stability assessment chart.

68


- if the factor of safety using this

clays, may be strain weakening. For these cases, post earthquake analysis should be carried out to consider the effects of earthquake induced strains on the available strength.

analysis is less than 1.0 using reasonable estimates of residual undrained strength, perform a post

liquefaction deformation analysis as described in 6.4.2.

There should not be an overeliance on stability 4. Should the results of the post

and deformation analysis, at the expense of

liquefaction analyses be unsatisfactory,

good engineering judgement, and the consideration of the general design principles

perform a real time coupled effective stress dynamic analysis. Both

given in Section 5.2. At best, the analysis

semi-coupled and fully coupled codes

methods are approximate, and controlled by the quality of data put into the analyses, and the limitations of the methods of analysis.

may be used for this purpose. Preferably, both pore pressure

generation and dissipation should be

7. ANALYSIS AND DESIGN

considered in the analysis. However, these analyses are expensive in computation and data required, and

OF CONCRETE DAMS

7.1 Past Performance of Concrete Dams in Earthquakes

they will only be needed for large dams, where the other analyses are giving marginal answers.

To date, concrete dams have performed well under earthquake conditions. No concrete

In all cases the extent to which one

analyses stability and deformation should be consistent with the hazard

dam has failed due to earthquake with loss of part or whole of the reservoir. However, a number of dams have suffered serious damage,

rating and size of the dam. In many

cases, it will be better to design

including cracking right through the concrete

remedial works, rather than continuing to do more and more sophisticated analysis.

eg. Koyna and Sefid Rud Dams. A summary of the earthquake effects on some concrete

dams is presented in Table 16. A number of these have been discussed in the literature (e.g.

Some soils (and rocks) in embankments

Clough and Ghanaat, 1994; Hinks and

and foundations, eg. overconsolidated Table 16. Earthquake Effects on some Concrete Dams

Dam

Height

Type

m

Koyna

Sefid Rud Pacoima

103 106 113

Arch

47

Curved gravity

Country

Concrete gravity India Iran Buttress

California,

Date

1967 Major cracking 1990 Major cracking 1971 Cracking at left

USA Lower Crystal Springs Blackbrook

29

Concrete and masonry gravity

Damage

Magnitude

M

6.5 7.3 to 7.7 6.6

abutment

California,

1906 No damage

8.3

USA UK

1957 Copings displaced.

5.3

Cracking

dam Hsingfengkiang Honen-Ike

105 30

Buttress

China

Multiple arch

Japan

1962 Major crack 1946 Crack in arch near

6.1

N/A

buttress

Ambiesta Maina di Sauris Shenwao

Redflag Rapel

59 136 53 35 110

Italy Italy Concrete gravity China

Arch Arch

Masonry gravity China Chile Arch

1976 1976 1975 1970 1985

No damage No damage Cracking Cracking, leakage Damage to spillway

6.5 6.5

N/A N/A 7.8

and intake tower

(from Hinks and Gosschalk, 1993)


69

Gosschalk, 1993;Serafim, 1981, Chopra and Chakrabarti, 1972, USCOLD, 1992). As an indication of the severity of shaking at some of the above dams, Pacoima Dam was subject to a peak ground acceleration of 1.2g in the

horizontal direction in the 1971 earthquake (Serafkn, 1981) and sustained minor damage. It suffered relatively minor damage during the Northridge Earthquake of 1994 despite peak

its foundation is not well known,

particularly important for foundatf gently dipping or near horizontally ! sedimentary rocks, or any rocks co

continuous gently dipping joints or sea 7.2 Defensive Design Measure

ground accelerations of 2.0g horizontal and

As for embankment dams, the use of d

1.4g vertical on the left abutment and 0.5g

measures for new dams, or for remedi on old dams is advised.

horizontal and vertical accelerations at the dam base. The water level in the dam was

41m below FSL at the time (Dames and Moore, 1994). Besides the dams in the above table, there have been many other dams subjected to severe earthquake loadings which have suffered no or very little damage at all (Serafim, 1981; Hinks and Gosschalk, 1993,

USCOLD, 1992). Of concrete dams generally, arch dams have been the best performers during earthquakes

These are followed by gravity dams. Buttress dams have not performed as well as the former

two types, probably due to the lower stiffness

The two main criteria for a concr

subjected to earthquake loading are remains serviceable after high pro4 earthquakes (e.g. the OBE) and t security of the storage is not prejudice to cause dangerous flooding after f probability earthquake (e.g. the MD„ ensure that these criteria are both sai there are a number of defensive measures which can be incorporated design of new dams or remedial work defensive design measures address dam structure and the interface of

with its foundations, and the foundation

in the cross-valley direction and the structural

In general terms, a dam should be d

discontinuity of the dam.

such that:-

While concrete dams have not generally

• sliding is prevented - or at least

suffered major damage from earthquakes, there is no cause for complacency. Many

older dams have had earthquake loads imposed on them which have been much greater than their design loadings (Hanson and Roehm, 1979). However, these dams have

generally been designed using a simplistic approach coupled with a low demand on the concrete tensile strength. Factors such as dynamic elastic modulus, interaction between the foundations, the dam and the reservoir,

resistance should be suffici prevent failure in the post ea condition after the MDE • the dam is stable against ove at all levels even after crac'

increased uplift pressure ind earthquake • any cracking will not I uncontrolled leakage • hydraulic outlet structures

considered. Now however, with dams being

damaged to the extent that th allow uncontrolled loss of wat might lead to collapse of the they cannot be used to lo

designed using more sophisticated methods

storage if necessary.

material strengths and stiffnesses under dynamic conditions were not normally

and with some dams relying on post-tensioned

ground anchors for stability, there may not be the same degree of built-in security. It is therefore necessary that the earthquake model adopted and the properties on which it is based for a particular dam and its foundations be well considered.

It is unrealistic to carry out a refined analysis if the structure and properties of the dam and

70

To achieve the above, the list belo some defensive measures that

implemented: • ensure that the cross section of a

well proportioned with no sudden in shape or section stiffness. • ensure that any superstructure on t' crest is minimal.


• ensure that there are no sudden changes in

the vicinity of the dam. The distribution of

the abutment profile which would give rise

this "added mass" was commonly based on Westergaard hydrodynamic pressure

to stress concentrations. • ensure that there are no geological features

in the foundations which would decrease the dam's sliding stability - if there are, then suitable means of stabilising these features or accounting for them in the

design should be employed. This is especially true of arch dams, since the interaction of dam thrusts and foundation characteristics is determinative of the whole dam-foundation stability. • ensure that there is sufficient internal and

foundation drainage, and that drain holes are of large enough diameter - the diameter

of drain holes is especially important if some controlled sliding is to be permitted. It is vital that the drains keep working or that the dam is stable post earthquake with the drains not operative. • if post-tensioned ground anchors are

distribution. At this time, the flexibility of the dam (especially in the case of gravity dams) and its dynamic response were not recognised. The next stage in designing concrete dams for

earthquake was to recognise the flexibility of the dam and linear elastic methods were established to determine a dam's response in the frequency domain (i.e. using an earthquake response spectrum). It is now becoming increasingly popular to determine a dam's response in the time domain (i.e. using

suitable accelerograms) when analysing the dam with non-linear elastic methods. The three main types of concrete dams are :

• gravity dams

required for stabilising a dam then ensure that they are unbonded (except for their anchorage length) - this will allow strains

• arch dams • buttress dams

due to any overload resulting from

Each type of dam puts particular demands on

earthquake loading to be taken over the entire free length of the cable rather than over a very short length either side of a

the numerical methods and the various parameters required for earthquake design and analysis. In these guidelines however,

crack in the dam or foundations - strains

emphasis will be given to concrete gravity

taken over the entire free length results in

dams because most concrete dams in Australia are concrete gravity structures, and arch and

much lesser strains and consequently much lesser stresses due to earthquake loading. • ensure that internal features such as

galleries do not give rise to stress concentrations which will lead to excessive cracking.

• provide so far as is practicable, sufficient outlet works discharge capacity to rapidly lower the storage if necessary, after an earthquake.

buttress dam analysis cannot be simply generalised. A review paper by Hall (1988) contains a good summary of the state of concrete dam seismic studies throughout the world. Over recent years, international dam

authorities have presented guidelines for earthquake analysis of concrete dams e.g.

USER (1977) and ICOLD (1986). There has 7.3.1 General

been considerable published work by Chopra (1979, 1987, 1991). A recent paper by Clough and Ghanaat (1993) presents an up to date

Until approximately forty years ago, the only

earthquake analysis and describes recent

concession made to earthquake effects in the design of concrete dams was to consider an

advances.

7.3 Analysis Methods

summary of the development of concrete dam

additional inertial load acting in the downstream direction. This load was detefmined by multiplying the mass of the dam by an assumed earthquake coefficient equivalent to an earthquake acceleration of 0.05g to O.lg. Later on, "added mass" was

applied to the upstream face of the dam to represent the dynamic effect of the water in

Methods of analysis range from simplified 2D methods such as presented by Fenves and Chopra (1987) to more complex, non-linear, 3D finite element, dynamic analysis. The type and degree of sophistication of the analysis method used for a particular dam will depend on a number of issues. These include

the size, type and hazard classification of the


71

dam, the valley configuration, foundation characteristics, the presence or otherwise of

storage siltation adjacent to the dam, and the characteristics of the earthquake.

The Fenves and Chopra (1986, 1987)1 for gravity dams uses a 2D analysis 1 based on modal superposition. It (ielcij the fundamental natural frequency and

In addition to the above issues there are a number of factors which need to be considered in assessing a concrete dam's ability to withstand an earthquake. These factors are concrete crushing strength, dynamic tensile

strength of lift surfaces, foundation strength and the amount of allowable sliding of the dam on its foundation for extreme

earthquakes. For all dams, foundation stability is paramount and in many circumstances it

will be more critical than maximum tensile stresses induced within the dam body. Relevant factors include the presence of weak seams, bedding surfaces and sheared joints and zones, joint orientations in relation to the thrust vectors imposed by the dam, continuity, roughness and shape, and the presence of low

strength and/or highly jointed rock. 7.3.2 General Description of Analysis Methods

(a) Type of Analysis

shape for the dam on a rigid foundalidn an empty storage. Empirically (U formulae based on a "standard" gravity cross section are used for this. 1 fundamental natural frequency is then ad to account for the interaction of the dan

the storage, the flexibility of the foundations, the compressibility of the^ in the storage and the effect of sedinic the bottom of the storage. 1 Kinj fundamental natural frequency correcte these factors and an appropriate euuh response spectrum, the spectral accele

(maximum acceleration of a single dcgi freedom (SDOF) oscillator) is delerini This is adjusted to account for the distril of mass and stiffness within the da compared to the equivalent SDOF oscil Knowing the distribution of accele through the dam, inertia and hydrodf forces for the fundamental mode a calculated. Formulae are provided to al

higher modes of vibration. The eart forces thus calculated can be include"

stability analysis or finite element ana' The increased power of personal computers and of finite element analysis programs means

determine instantaneous stresses in the

that much more sophisticated numerical

The above methods would normal appropriate when cracking of the dam

models of concrete dams can be used to determine the dams' behaviour during earthquakes. However, not all concrete dams

will need dynamic, numerical analysis having a high level of complexity. There is still a need to carry out simple analyses without having to resort to finite element analysis. For

example, a small (less than 15m high) concrete gravity dam with a length more than twice the height can be quite adequately examined using simple, pseudo-static methods. In this case,

the dam is considered to be a rigid body on rigid foundations with an added "virtual" mass of water attached to the upstream face. An

inertial load due to the earthquake is included in the traditional, cantilever stability analysis (see the ANCOLD "Guidelines on Design Criteria for Concrete Gravity Dams",

ANCOLD, 1991). For higher concrete gravity dams which are still essentially two dimensional structures, the simplified dynamic analysis methods of Chopra and others are appropriate, especially in preliminary analysis and design.

72

expected from an earthquake.

With the Fenves and Chopra (1986, method, appropriate parameters (repr the influences of dam/foundation inte dam/storage interaction, mitigating effl silt, higher vibration modes) need/ chosen. If there is sediment at the bo the storage then there will be some abs of the hydrodynamic wave. This will t' a reduction of the lateral forces on the the dam. The paper gives some typical ,

The Fenves and Chopra method uses frequencies and mode shapes basedj triangular cross section dam. For i accurate estimation of natural frequenci

mode shapes, especially if the dam's? section is more trapezoidal or has a Vj slope on the downstream face, a 2 element model can be used. ^

When cracking of the dam due to co static and dynamic loads is expecte


linear methods can be used. While linear

would normally be considered, only for final

elastic models can be analysed in either the

design work where the high cost of the

frequency (using a response spectrum), or time domain (using an accelerogram), a non-linear

analysis is an economic proposition when compared to the total project cost.

analysis would normally be analysed in the time domain. This type of analysis would normally require a quite sophisticated finite

(b) Dimension of Analysis

element program. However, an approximate

analysis can be made using the approach of the US. Corps of Engineers as described by Guthrie (1986) which is based on a linear

Whether to do a 2D or a 3D analysis will largely depend on the type of dam and the valley geometry:

elastic analysis. Larger damping factors are

• Arch dams will certainly require a 3D

used to simulate the energy absorbing effects

analysis. • Buttress dams, especially those with buttresses having low stiffness in the cross-

of cracking. For arch dams, buttress dams and gravity dams

where length is less than twice the height, a three dimensional finite element model can provide natural frequencies and mode shapes

for determining earthquake loads. As the mode shapes are more complex than those for

a gravity dam analysed in two dimensions, it is more appropriate to analyse these dams in the time domain.

valley direction, will also require a 3D analysis.

• Gravity dams in wide valleys may be examined using a 2D analysis. This is especially so if the ends of the dam are relatively unrestrained (e.g. earth embankments at the ends of the concrete dam). • Gravity dams in narrow valleys where the

Where cracking is expected in arch dams,

length of the dam is less than twice the height of the dam should be examined

buttress dams and gravity dams with lengths

using a 3D analysis. However, 2D analysis

less than twice their heights, non-linear analysis is complex. A non-linear analysis

may be used for preliminary design. Table 17 summarises this advice.

Table 17. Analysis Methods Method

Dimension 2-D

Pseudo-static

Simplified Dynamic Analysis or linear elastic finite element method US Corps of Engineers

2-D 2-D

(as described by Guthrie (1986)) Linear elastic finite element method

3-D

Non-linear Dynamic Analysis

3-D

Applicability Gravity Dams < 15m high Length > 2*Height Gravity Dams - No cracking

Length > 2*Height Gravity Dams - with cracking Length > 2*Height Gravity Dams - Length < 2*Height Arch Dams Buttress Dams

Gravity Dams - Length < 2*Height Arch Dams Buttress Dams

(c) Foundation Characteristics Unless the foundation rock is considerably stiffer than the concrete in the dam which would oe unusual, it will be necessary to

reduced and the effective damping is increased. If the foundations are included in a finite element model, they are given zero mass, but the modulus and damping coefficient of the rockmass is included.

include the effects of the dam-foundation interaction. As the foundation rock becomes

(d) Hydrodynamic Pressure

less stiff in comparison with the concrete in the dam, the natural frequency of the dam is


73

The water pressure on a dam is increased

compression are discussed in S

during an earthquake by an added hydrodynamic pressure. The hydrodynamic

7.6.1.

pressure results from the dynamic interaction of the dam and the stored water. The lateral forces on the dam resulting from hydrodynamic pressure are mostly determined by the slope of the dam's upstream face, the

(g) Analysis Program A number of the papers referred to i may refer specifically to a progra EAGD-84 for 2-D analyses and BAG

flexibility of the dam and the amount of

3-D analyses.

sediment on the bottom of the storage. The

storage water affects the dynamic behaviour of the dam which in turn affects the hydrodynamic pressure on the dam.

These programs were written by C association with others, and are availa

the Earthquake Engineering Research University of California, Berkeley.

Hydrodynamic pressure is discussed in more detail in Sub-section 7.4.2.

Program EAGD-84 is user frien includes the effects of: 1

(e) Non-linear Behaviour and Cracking When the combined maximum static and

• the reservoir; • sedimentation at the reservoir botto

dynamic tensile stress in a concrete dam

• water compressibility; and

exceeds the dynamic tensile strength of the concrete (especially at lift surfaces) or of the

• dam-foundation rock interaction.

foundation interface, cracking and non-linear behaviour can be expected. The cracking and non-linear behaviour will reduce the stiffness

Input into the program include horizontal and vertical accelerati histories. The difficulty here is in

of the dam and provide a mechanism for

locally recorded accelerograms upw| represent an accelerogram for say a

energy absorption. The peak values of tensile stress and extent of tensile zones will tend to reduce.

If extreme earthquakes are permitted to cause

significant cracking of the dam then consideration has to be given to additional factors:

(i) what uplift pressures are induced in

Design Earthquake. This is discus Section 7.4. A suitable plot of resul different accelerograms will generally it whether the sensitivity range is too grea 7.3.3 Details of Analysis Methods (a) Pseudo-static Method

the cracked zone?

(ii) how is crack propagation modelled? (iii) what are the limit states for the dam?

This method assumes that the dam foundations are rigid and that the wate

Point (i) is discussed in more detail in Sub¬

storage is incompressible. It assumes f, entire dam has an acceleration the sam peak ground acceleration.

section 7.4.3.

On point (ii), crack propagation may be modelled using zero tensile strength finite elements (i.e. cracking is distributed throughout elements, modelling discrete cracks where there are tensile stresses, or using a fracture mechanics approach).

The hydrodynamic pressure distribute calculated according to Westergaard (1 Zangar (1952) e.g.

Ph = C—wh

g

Point (iii) is discussed in Sub-section 7.3.4. (1) Dynamic Strengths Dynamic strengths and modulus of concrete and the rock foundation in tension.

74


Ph = Hydrodynamic pressure ag = Peak ground acceleration g = Acceleration due to Gravity

w = Unit weight of water C = Dimension less coefficient:

C=

• earthquake - peak ground acceleration - response spectrum for the peak ground acceleration and a range of damping ratios • dam/foundation/storage interface

^£(22h

y = Depth from storage surface to particular level h = Total depth of reservoir at Section Cm = Maximum value of C (0 73 for Vertical Face)

- period lengthening ratio and added damping ratio due to hydrodynamic effects - period lengthening ratio and added damping ratio due to dam - foundation rock interaction

The force derived from this pressure

distribution, and the inertia force equal to the mass of the dam multiplied by the peak ground acceleration, are then included in the

traditional cantilever stability analysis with the

- hydrodynamic pressure distribution (depending on the degree of hydrodynamic wave absorption in the alluvium and sediments at the bottom of the storage).

appropriate static forces. The computational steps for the method are:

The stability analysis gives peak stresses and sliding stability for the dam. However, it should be remembered that the resulting peak stresses only act for a very short time during a cycle of the earthquake. Consequently, peak stresses greater than dynamic strength may be

indicative of zones of potential cracking. Analysis of the dam in a post earthquake cracked condition will be required to assess the overall stability. Generally, this method would only be used as a screening method to determine if a gravity dam has a potential problem with earthquakes or for small dams less than 15m high, in a

1. Compute the fundamental period of the dam for an empty reservoir and

rigid foundations. For dams of triangular cross section, use the formula:

Ti = 0.38

Ti =

Hs VEs"

where:

Fundamental period for dam of essentially triangular cross

section with an empty

wide vallty.

reservoir and a rigid

(b) Linear Elastic Dynamic Methods Hs =

(i) Simplified Dynamic Analysis Method

Eo =

foundation Dam Height in metres Dynamic Elastic Modulus of Concrete in MPa.

This method is presented by Fenves and Chopra (1986, 1987). This sub-section will describe the method in general terms but Fenves and Chopra (1986) should be read for more details.

The input parameters required fall into three

2. Adjust the fundamental period using the period lengthening ratios to account for the influence of the impounded water , the flexibility of the foundations and for the presence of sediments at the bottom of the reservoir.

groups -

• dam geometry and properties - height of dam - modulus of elasticity of concrete - the generalised mass of the dam - the generalised earthquake force

coefficient for the dam

3. Compute the ratio of the fundamental vibration period of the impounded water to the fundamental period of the dam.

4. Compute the damping ratio for the dam using the basic viscous damping ratio for the dam on rigid foundations


75

and empty reservoir (say 5%) and the added damping ratios accounting for the extra damping due to the dam

to the static forces present pri(

foundation rock interaction and

The above calculations can convene done on a one page spreadsheet

hydrodynamic effects. 5. Determine the hydrodynamic pressure

earthquake in a stability analys

Figures in Fenves and Chopr;

distribution using the period ratio

1987) show the hydrot

calculated in step 3.

pressure profiles required for ai

6. Determine the fundamental mode

A parametric study of a large concrete

shape for the dam and compute the generalised mass (Mj) and earthquake coefficient (Lj). Conservatively, Lj/ Mj = 4 for dams with full storages and Lj/ Mj = 3 for

dam was carried out by Chavez and (1993) using the above method. The r< the study indicates the sensitivity of t

dams with empty storages.

(ii) US Corps of Engineers Method

7. Compute the equivalent lateral earthquake forces distributed through the dam for the fundamental mode, using:

sliding to the various parameters.

This method is as presented by Guthrie It can be used for both 2D and 3D prob For 2D problems, the Simplified E Analysis above can be used to de earthquake forces using a viscous d

^ (y) = Sa^"^ [(ds (y)Ky) + gPi (y.T r)] Mi g

ratio for the dam of 5%. The forces i applied as equivalent static loads togetl other static loads to a 2D finite elemen of the dam and foundations. Peak stres

where :

computed using the finite element modi f, (jy) = Lateral earthquake force at elevation y Li = Generalised earthquake

coefficient Mi ::= Generalised Mass ¦SaCf,,^,) " Spectral acceleration for period

7, and damping^ g = 9.806 m/s^

fi)s(y) = Weight of dam per unit height Tr) ~ Hydrodynamic pressure for the

adjusted fundamental period of the dam 8 Compute earthquake forces due to higher vibration modes. 9. Compute the square root of the sum of

the squares of the forces calculated in steps 7 and 8. Add the resulting forces

If the dam has to be considered a? problem then a full dynamic analysis i 3D finite element model has to be cari to determine peak stresses for the 5% 1 damping. For a more detailed descrip the full dynamic analysis see below. The analysis is carried out for be operating basis earthquake and the ma design or credible earthquake. It cQ that the dam behaves elastically for stresses less than 10% of the c compressive strength. For tensile i

beyond that value the concrete will craq method assumes that even though the d? now behave non-linearly, it can s

modelled elastically provided the di ratio is increased. The increase in di ratio is assumed to represent the in<

energy dissipation caused by tensile en The steps in the method are best repn by flow charts which are taken from < (1986). However in general terms, the I follows two streams - one for the ma

credible earthquake (or maximum earthquake) and one for the operatinj earthquake:

-4

76


Figure 33 Sequence of Analysis for Maximum Design Earthquake

(Guthrie 1986)


CONTINUED

Figure 33 (continued) Sequence of Analysis for Maximum Design Earthquake

(Guthrie 1986)


Maximum Design Earthquake (refer Figure 33)(i) Initial 2D analysis, e.g. as described by Fenves and

Chopra (1986, 1987) with 5% viscous damping.

horizontal plane. Refer to the sub-section below on Permanent Deformations for a

suitable method. If the permanent displacement is

considered acceptable then check post earthquake stability for maximum static loading,

(ii) If peak tensile stress (static +

FEM (good quality lift

considering the cracking caused by the earthquake and the extra uplift force that might be caused by that

surfaces assumed). If peak

cracking. If the permanent

tensile stress (static +

dynamic) <15% UCS of

displacement is considered unacceptable then the dam is

concrete then the dam is

unsatisfactory and

adequate for MDE.

strengthening (or some other remedial option) for earthquake loading is

dynamic) >15% UCS (fc) of concrete then re-model using

(iii) If peak tensile stress (static +

dynamic) is >15% but <20%

required.

UCS of concrete then repeat

FEA with 7% Damping Ratio.

Operating Basis Earthquake (refer Figure 34) -

Assume cracking wherever

combined tensile stress (static

+ dynamic) is >10% UCS of concrete.

(iv) If peak tensile stress is >20% UCS of concrete carry out (iii) but use a 10% Damping Ratio. (v) On horizontal planes with cracking, carry out sliding analysis with cohesion only applied to the uncracked portion.

(vi) If sliding safety factor (factored sliding strength divided by horizontal forces causing sliding) >1.0 then check for post earthquake stability. If sliding safety

(i) If the above analysis indicates no cracking then further analysis for the operating basis earthquake is not required.

(ii) If analysis for the operating basis earthquake is required then carry out an analysis

similar to (i) for the maximum credible earthquake using 5% damping ratio. The peak tensile stress (static + dynamic) is not to be greater than 10% UCS concrete provided the lift joints are sound.

Fenves (1989) compared the analysis of a gravity dam using the above method with

factor >1.3 and maximum compressive stress <0.5 UCS, concrete dam is satisfactory.

more rigorous non-linear methods. His conclusions were that assessing the extent of

If either these two criteria are not satisfied then the dam is unsatisfactory and strengthening (or some other / remedial option) for earthquake loading is

give reasonable accuracy for the five cases of the one dam analysed. He warned however, of

required.

(vii) If sliding safety factor <1.0 then compute the displacement along a

cracking from the linear analysis appeared to

extrapolating the generality of the method. This is due in part to the arbitrary levels of maximum tensile stress and viscous damping ratio. He finishes by saying that "there is no theoretical basis for assuming that tensile cracking results in the increased energy dissipation implied by greater damping ratios."


79

Figure 34 Sequence of Analysis for Operating Basis Earthquake

(Guthrie 1986) The preceding method could probably be used

A non-linear dynamic analysis of a concrete

for arch and buttress dams as well. However,

dam is a complex analysis and would normally be undertaken by specialist numerical analysts experienced in such work. It would normally be done only for major dams where the cost of the new dam or the cost of remedial works for

consideration would have to be given to the appropriateness of the damping ratios and corresponding limiting tensile strengths used. (c) Non-linear Dynamic Analysis

80


an existing dam is sufficient to justify the greater expense of this type of analysis.

Provided peak tensile stresses (static + dynamic) do not cause cracking in the dam,

then a linear analysis will suffice. When cracking occurs, stresses will re-distribute. Therefore, when cracking is expected or there are pre-existing cracks (e.g. open vertical

contraction joints) and the dam is essentially 3D m nature then a non-linear analysis should be done.

There are three main ways in which cracks can

be modelled: • as distributed (smeared) cracks (zero elastic modulus in direction perpendicular to cracks) • as discrete cracks at finite element interfaces • using fracture mechanics.

The first of the alternatives is computationally the simplest. Further details on cracking are

given by Zienkiewicz, Valliappan and King

(1968), Rashid (1968), Mohraz, Schnobrich and Gomez (1970), Darwin and Pecknold

(1978), Phillips and Zienkiewicz (1976), Bazant and Cedolin (1979), Bazant and Ob

(1979), Argyris, Krempl and William (1977), Gerstle (1981), Kotsovos and Newman (1978), William and Warnke (1975), Cedolin, Crutzon and Dei Poli (1977), Bicanic and Zienkiewicz (1983), Zienkiewicz, Fejzo and Bicanic (1983), Zienkiewicz, Hinton, Bicanic and

Fejze (1980), Pande and Shen (1982), Pal (1974) and Chapuis, Rebora and Zimmermann

(1985). A non-linear analysis will by its nature, require a time-history analysis. Consideration

will have to be given to: • the way the compressibility of the storage water is modelled • the way seepage pressures particularly

those due to water penetrating cracked zones, are modelled. The non-linear analysis of concrete dams is

still a developing and specialised field. Other relevant papers include Waggoner, Plizzari and Saouma (1993), Gao Lin, Jing Zhou and Chuiyi Fan (1993), Greeves and Taylor (1992), Cervera, Oliver and Galindo (1992),

Jing Zhou and Gao Lin (1992), Clough and Ghanaat (1993), Fenves and Mojtahedi (1993).

7.3.4 Analysis of Permanent Deformations In some concrete gravity dams subjected to

say the maximum credible earthquake, it may be permissible for the dam to slide on its base or within the foundations and be permanently deformed after the earthquake. This of course assumes that during or after the deforming process, the security of the storage is not

prejudiced allowing for the potentially increased uplift pressures and lower strength which may apply. Researchers such as Chopra

and Zhang (1991), Leger and Katsouli (1989) and Danay and Adeghe (1993) discuss calculations which indicate that typical permanent displacements for large concrete

gravity dams subjected to earthquakes having peak ground accelerations the order of 0.5g can range from tens of centimetres to more than half a metre. However, some dams with

suitable foundations would be able to withstand small displacements. Dams relying on post-tensioned ground anchors or drain

holes for normal load static stability, might not be able to withstand these sort of movements

if the displacement was sufficient to shear the anchors. However it may be acceptable to

have the anchors sheared for a low probability earthquake provided the dam was stable under the post earthquake load case.

The type of analysis required to compute permanent deformations is similar to that carried out for embankment dams i.e. Makdisi

and Seed (1978). The dam is considered as a block with a limiting sliding strength along its foundations. The dam is subjected to a time varying input of acceleration. When the acceleration is greater than the limiting acceleration (the acceleration causing inertia forces which are greater than the sliding strength of the foundations) the dam will move on its foundations. The parts of the accelerogram greater than the limiting acceleration are double integrated to obtain cumulative displacements.

The type of analysis just described is a requirement of the US Corps of Engineers method for dynamic analysis of concrete gravity dams when the sliding safety factor is less than one. The sliding analysis is carried out for horizontal planes through the dam where there is cracking. A crack is assumed

through the dam with suitable slip elements along the crack interface.


81

T! In their study of base sliding, Chavez and Fenves (1993) found that sliding was not likely to occur if the storage is less than half

The response spectrum used in the analysis of

a dam should be site specific and relate to the

full. Other conclusions were :

peak ground accelerations examined. The response spectrum should also reflect the

• vertical ground motion has almost no effect

frequency mix and duration of the design earthquakes. It will therefore be derived from

on the sliding displacement (it slightly

a number of earthquakes having various

increased the maximum stresses). This is

epicentral distances from the site and

partly because the vertical and horizontal

consequently, different acceleration attenuation functions. The response spectra

ground motion do not necessarily coincide.

• the assumption of rigid foundation rock can significantly overestimate the amount of base sliding

should be obtained from a seismologist as part of the assessment of seismicity of the dam site.

(b) Accelerograms At the present stage of development of the computation of permanent deformations and

Where a time-history analysis is to be done, at

the determination of suitable maximum displacements there is still much work to be done especially regarding local conditions. These guidelines therefore counsel that considerable care be used if a dam is to be allowed a permanent deformation following an earthquake. Adequate sliding and overturning stability must exist after the earthquake using foundation strengths appropriate to the displacement (usually the residual strength) and uplift appropriate to the displaced condition, allowing for opening of joints and bedding, and for reduced (or no) drainage

least three different accelerograms appropriate to the dam site and for a particular peak ground acceleration, should be used. These accelerograms may be recorded accelerograms

which are suitably scaled (accounting for change in frequency mix and phase with change in peak ground acceleration) or synthetic accelerograms which fit the response spectra for the site. Care should be taken in

selecting accelerograms which are similar to Australian earthquake conditions, and advice should be obtained from a seismologist.

capacity.

7.4.2 Hydrodynamic Pressures

7.4 Design Earthquake and Hydrodynamic Loads

Any movement of the dam and foundation will

7.4.1 Earthquake Parameters

As discussed elsewhere in these guidelines, dynamic analysis can be carried out in the frequency domain or the time domain. For the former, response spectra are required while for the latter, accelerograms are required.

cause movement in the water of the storage and in turn, the pressures generated by the water will impose forces on the dam.

Engineers have traditionally used hydrodynamic pressures derived by Westergaard (1933). These pressures are commonly converted into equivalent lumped 'virtual' masses which are attached to the dam. Westergaard's pressure distribution assumes

(a) Response Spectra

that the water in the storage is incompressible and that the dam and its foundations are rigid.

A response spectrum shows the extent to

However, this is not always so. In high

which any single degree of freedom structure with an assumed level of damping would

gravity dams and slender arch dams especially, where the dam is flexible, there can be considerable interaction or coupling between

respond to particular earthquakes.

the dam and the storage.

Knowing the natural frequencies of vibration and the corresponding mode shape for a structure, the spectral accelerations

corresponding to particular natural frequencies and damping ratios can be converted to inertial loads.

Considerable work on the interaction of gravity dams and their storages has been done by Chopra and his fellow researchers. Details of the work are given in Chopra (1967), Chakrabarti and Chopra (1974), Chopra,

Chakrabarti and Gupta (1980), Chopra and Gupta (1981). Other relevant papers include

82


Clough and Chang (1980), Dungar (1978),

Guthrie (1986) uses the pre-crack uplift

Hall (1988), Zienkiewicz, Paul and Hinton

pressure diagram.

(1983) and Tsai and Lee (1989). These guidelines recommend that for the Besides the lumped, 'virtual' mass approach,

duration of the earthquake, the pre-earthquake

the interaction of a dam with the water in its storage can be determined by treating the

uplift pressure distribution is used for the stability analysis. However for the post earthquake analysis, consideration should be

water as a 'solid' having zero shear modulus

but retaining compressibility. This approach, although simple in principle, has a number of

given to the amount of cracking and the post-

numerical problems. An alternative approach

system. In the post-earthquake situation, full headwater pressure is assumed to exist in a

is therefore preferable where from the beginning, the shear components of stress in

the fluid are neglected. This latter approach includes the effect of water compressibility. It also will allow the limiting case of incompressible water to be considered.

Accounting for water compressibility and coupling of a dam with the water in its storage adds considerable computational effort to determining the effects of earthquake loading

earthquake efficiency of the dam's drainage

crack at least as far as the line of drains. If the drains have sufficient capacity and they have not been disrupted by sliding of the dam, then, if the crack extends, past the drains, a

significant reduction in uplift pressure should be considered. Typically, the pressure at the line of drains in this case might be the tailwater pressure plus 50% of the difference between headwater and tailwater pressures.

on a dam. Neglecting coupling and water

The pre-earthquake uplift pressure distribution might have had a 67% reduction. The lesser

compressibility in the simpler "added mass" approach is not considered significant for

for the greater amount of drainage with which

excitations at frequencies below the natural frequency of the reservoir.

An estimate for the fundamental frequency of a reservoir can be obtained from:

f =_£_ w 4H

eff

where: fw = the fundamental frequency C = the compression wave speed

in water (l,439m/s) Heff = an effective depth of the reservoir.

The above relationship was obtained from Duron, Ostrom and Aagaard (1994) and applies strictly to an infinite reservoir of constant cross section. Duron and Hall (1988) indicate that if the ratio of to the

reduction for the post-earthquake case allows the drains would have to cope.

7.5 Design Criteria 7.5.1 General Approach

The working group has had two major difficulties in preparing suitable design criteria for concrete gravity dams.

(1) The current ANCOLD (1991) guidelines for design criteria for concrete gravity dams are based on a limit state approach with partial factors of safety. This method has proven to be difficult to use on dams subject to significant earthquake loads.

The ANCOLD (1991) guideline is under review to address these problems. In the

fundamental frequency of the dam-foundation

interim, it is recommended that the design loadings and acceptance criteria described

alone (fj) is near unity, water compressibility will have a significant effect. For ratios of fw

the BC Hydro (1995) guidelines.

herein are used. These are based largely on

to fj much greater than one (e.g. >1.5)

incompressible fluid behaviour can be

(2) Existing guidelines for design of

assumed.

concrete gravity dams are not simply applied to a risk based approach. As a result it has not been practicable to develop these guidelines to directly apply to a risk based approach, and they are given in terms of a deterministic method using OBE and MDE. For

7.4. i Uplift/Seepage Pressures

The USBR (1977) considers that the uplift pressure within the crack is zero while ICOLD (1986) assumes full headwater pressure but recognises the need for further research.

completeness, flood and static load case are

also listed. Those wishing to use a risk based


83

(D + H +1 + S + U)

approach may do so, taking account of the

general intention of the load cases detailed herein.

(b) Unusual (Flood) Load Case

7.5.2 Loads and Load Cases

Permanent and operating loads of the Usual Load Case, except for ice loading, should be

The loads considered in the assessment of concrete gravity dams and their foundations

should include the following:

considered in conjunction with reservoir and

tailwater levels and uplift resulting from the passage of the IDF.

(D + H' + S + UF)

• Dead loads of permanent structures,

equipment and foundation rock (D). • Water load due to maximum normal

headwater level combined with the most critical concurrent tailwater level (H). • Water load due to maximum flood headwater level based on the Inflow Design Flood (IDF) with corresponding tailwater levels (H). • Foundation uplift (U), both at the concreterock contact and at critical discontinuities within the foundation. • Static and dynamic thrust created by an ice sheet, for reservoirs subject to freezing (I).

• Vertical and horizontal loading due to rock or soil backfill (both natural or engineered), including potential effects of liquefaction and loads from silt deposited against the dam (S). • Load due to Operating Basis Earthquake

(OBE) Q' • Load due to Maximum Design Earthquake

(MDE) (Q).

The potential should also be assessed for reservoir levels higher than would result from passage of the IDF, such as those due to operating failures or other unusual conditions.

The effects of ice loads should not be considered simultaneously with flood conditions.

(c) Unusual (Earthquake) Load Case Permanent and operating loads of the usual

load case except for ice loading, should be considered in conjunction with earthquake loading associated with the Operating Basis Earthquake (OBE). The effect of ice loads should not be considered simultaneously with OBE earthquake conditions. The analysis should be carried out for the dam

Determination of the loads should take into account the actual field conditions and instrumentation records.

Foundation uplift assumptions should reflect the stress state and condition of the dam and foundation. Disruption of the dam and/or foundation condition due to an earthquake should be recognised in assessing the uplift assumptions for the post-earthquake case.

The dam and foundation should be assessed for the following load cases: (a) Usual Load Case permanent and operating loads should be considered for both summer and winter conditions including self-weight, ice (where applicable), silt, earth pressure, and the maximum normal reservoir level with

appropriate uplift pressures and tailwater level.

84

where subscript "F" refers to the flood case.

empty case.

(d) Extreme Load Case Permanent and operating loads of the Usual Load Case should be considered in conjunction with seismic loads of the Maximum Design Earthquake (MDE).

(D + H + S + Q + U) The effects of ice should be given special consideration, recognising the high uncertainty associated with ice loading on earthquake loading, and its effects on the dam. (e) Other Load Cases Where earthquake-induced cracking at the concrete-rock interface or any weak section is

identified, a stability analysis should be carried out to assess whether the dam in its post-earthquake condition is capable of resisting loads of the Usual Load Case.


have been identified. Combinations with other loads would be site specific.

(D + H + S + UpQ) where subscript "pg" refers to the postearthquake case.

An inoperative drain case assuming plugged drains may be assessed and could be considered as an Unusual Load Case.

Concurrent ice loading (with the postearthquake condition) may be considered in areas where appropriate.

7.5.3 Acceptance Criteria

A landslide generated wave case should be considered where existing or potential landslides, which may affect the reservoir,

All kinematically feasible failure modes, analysed by the single-slice, rigid-body, force equilibrium method, should satisfy the acceptance criteria shown in Table 18.

Table 18 Stability Index Acceptance Criteria Load

Sliding Factor

Position of Resultant Force

Minimum Compressive

(Note 1)

(Note 2)

Stress Factor

(Note 3) Usual

1.5-2.0

Mid-third of surface

4.0

No tension

Unusual (Flood)

1.3-1.5

and

Mid-half of surface

2.7

One-quarter tension

Unusual (OBE) Extreme (MDE)

1.1 - 1.3

Within surface

1.3

Post-earthquake

1.2-1.4

Mid-half of surface

2.7

One-quarter tension Notes: 1. Sliding Factor (Frictional Analysis) = Resisting Forces

Applied Force Lower values of the range apply where the geology and the strength parameters are reasonably well known. 2. Vector summation of all forces, including uplift, acting on the analysis surface 3. Compressive Stress Factor = Unconfined Compressive Strength Compressive Stress Normal to Surface To be considered primarily for massive but low strength rock and weak deteriorated concrete.

7.5.4 Post Earthquake Stability

pressure distribution should be as discussed in sub-section 7.4.3, i.e. full headwater pressure

If a dam is likely to be severely damaged after being subjected to the MDE, considerable time may elapse before the dam can be repaired or the storage lowered. Consequently, all parts of

the dam will need to remain stable after an extreme earthquake event. The stability of the dam should therefore be checked for static loading conditions. The assumed uplift

within cracks emanating from the upstream face.

7.5.5 Foundation Stability Where there is the possibility of a sliding failure along faults, shears and/or joints, the stability of the foundations should be


85

examined. This examination should be made for both the earthquake and post-earthquake cases. The dam itself should be examined for local overstressing due to foundation deficiencies.

recognised that a single spike of excessive

localised tension should not be taken to represent dam failure. In consideration of the above however, these

guidelines recommend that for sound lift Sliding failure is especially likely when

surfaces, the apparent tensile strength to be

discontinuities and/or horizontal or subhorizontal seams close to the foundation surface contain clay, or have been previously sheared eg. bedding surface shears due to stress relief, folding, or associated with faults.

used is 16% of the standard compressive strength.

7.6 Dynamic Material Properties

For dynamic modulus of elasticity, Clough and Ghanaat (1993) suggest a value 25% greater than the static value and these guidelines recommend this be adopted. For existing

7.6.1 Concrete

may be determined using geophysical means

The compressive and tensile strengths of

velocity). Values obtained should be compared with static and dynamic small sample laboratory test values for credibility.

dams, the elastic modulus of the concrete mass (e.g. derived from measured shear wave

concrete increase with increased rate of

loading. The dynamic compressive and tensile strengths of concrete can therefore be expected to be greater than the static strengths. As dynamic compressive stresses are rarely of concern, the allowable compressive stress for

static loading can be used also for dynamic loading.

Raphael (1984) states that the apparent tensile strength of concrete under seismic loading

which should be used with linear finite element analyses is given by:

7.6.2 Rock

In most cases the stability of the dam will be controlled by sliding in or on the dam rock foundation. To carry out static and dynamic analyses, it will be necessary to: • assess and map surface exposure,

excavations and drill core to define the geology of the site. In particular the 3dimensional orientation, continuity and

fr = 0.65 fc 2P

detailed nature of bedding, joints, and other features such as shears are required.

where fc is the concrete compressive strength

in MPa and fr is the apparent seismic tensile strength in MPa. Values given by this formula are some 50% greater than the apparent tensile

strength for static loading. Raphael suggests that fr is twice the splitting strength of the

• the shear strength of the foundation rock should be determined using appropriate rock mechanics techniques, such as those

described in Hoek (1983, 1990, 1994), Hoek and Brown (1980), Patton (1966),

concrete under static loading.

Barton and Choubey (1977), Barton and Bandis (1991). These require an

Clough and Ghanaat (1993) suggest that the

assessment of the orientation, spacing, continuity, shape and roughness of

apparent dynamic tensile strength is about 25% greater than the measured static value which gives apparent tensile strength about 20% of the standard compressive strength.

They further suggest that there may be a 15 to 20% loss of strength across lift joints. These figures may be even lower for poorly

discontinuities in the rock (e.g. joints), and the strength of the rock substance as this varies with confining stress. Care should be taken in applying these techniques, to

fiowever, the peak dynamic tensile stresses

account for the presence of continuous, adversely oriented low strength surfaces such as bedding surface shears, faults or shears, and to take account of the mechanisms of failure.

only exist during a fraction of a response cycle. Even though these peak stresses may

The Hoek, and Hoek and Brown methods give

constructed or defective lift surfaces,

greatly exceed the tensile strength of the concrete, any cracking that might be initiated will not have time to fully develop. It is well

86

strengths for "undisturbed rock" and

"disturbed rock". The latter should generally be adopted unless advice from an experienced


undisturbed rock strengths apply to confined

there is proper access to these structures. Bridges and roads may need to remain in a sound state after an earthquake depending on their importance.

conditions such as underground openings, where dilation or shearing causes increases in

Generally, appurtenant structures should be

normal stress.

such that:

It will be noted these stress methods give non linear failure envelopes, with high friction

• they maintain their normal operating

rock mechanics specialist indicates otherwise, because of the uncertainty as to the accuracy of the Hoek methods, and since the

angle and low cohesion at the normal stresses

condition after an operating basis earthquake

usually applicable to dams. They also allow

• they are not damaged to an extent where

estimation of the modulus of the rock mass

they could allow sudden or uncontrolled

(Hoek, 1994). The dynamic modulus may be higher than the static modulus as discussed in Clough and Ghanaat (1993) and Scott and Von Thun (1993) and may, as for concrete, be obtained by geophysical means, or by relation

loss of water from the storage for a more extreme earthquake up to the maximum design earthquake.

8.2 Performance Requirements

to the static modulus.

A dam's foundations will normally contain joints, shears, and bedding. Consequently, it

will not be possible to transmit tensile stress within the foundations and the allowable tensile strength for the foundations will therefore normally be assumed to be zero. However, if extensive site investigation and

This sub-section gives the performance criteria

for the operating basis earthquake (OBE) and the maximum design earthquake (MDE) which could be the maximum credible earthquake (MCE). Performance requirements are given for a number of appurtenant structures including intake towers, outlet conduits, outlet

strength testing is able to prove that the

works, spillway gates, spillway piers, spillway

foundations for a particular dam site are capable of transferring tensile stresses, then

access roads.

the tensile strength of the rock may be included. Where foundation rock strength becomes

gate hoist piers, access bridges and piers,

Intake Tower

OBE: Static and dynamic loads to induce maximum concrete and steel reinforcement stresses

critical, as they often will, advice should be obtained from a person expert in rock mechanics.

which satisfy AS3600

8. APPURTENANT

reinforcement yielding) and

(Concrete Structures Code) (i.e. limited amount of

STRUCTURES

the tower and its base remain stable.

8.1 Introduction A number of subsidiary structures associated with a dam are essential for the dam's operation. Consequently damage to or destruction of these appurtenant structures

MDE: Significant amount of reinforcement can yield horizontal reinforcement

designed to prevent vertical reinforcement from buckling and to contain concrete when it is in compression (i.e. concrete contained between inner and outer layers of

would be prejudicial to the dam's safety. An important facility at a dam is one that allows water to be released in a controlled manner. If

therie has been an earthquake and the dam is damaged to the extent that the dam is not serviceable then it may be necessary to lower the storage so that remedial works can be undertaken. It will therefore be necessary that not only the outlet structures and their gates and valves remain serviceable but also that

vertical reinforcement will not spall away). Outlet Conduit


87

OBE:

Static and dynamic loads to

loads should satisfy ASS 600

induce maximum concrete and steel reinforcement stresses

(Concrete Structures Code).

Earthquake loads from the orthogonal directions may be

which satisfy AS3600

combined on a square root of the sum of the squares basis.


Dynamic loads to include those induced from the earthquake effect on the

MDE: Carry out dynamic analysis as for the OBE. Combined static

overlying dam.

and dynamic loads may cause

MDE: Conduit not to collapse or

cracking but piers must

rupture - Collapse could lead to undermining and subsequent failure of an

remain stable for overturning

and sliding. Piers should not be permanently displaced to the extent where the spillway

overlying embankment.

Rupture could cause piping or destabilise an embankment by a marked increase in pore

gates become jammed.

Spillway Gate Hoist Piers

pressure.

OBE: Carry out appropriate dynamic analysis similar to that for spillway piers. Combined static and dynamic loads

Outlet Works

OBE: All valves to maintain their

should satisfy AS3600

normal operating capabilities.

(Concrete Structures Code). MDE: Emergency closure and

MDE: If the gates can be operated from an alternative position (possibly, in a less efficient manner), then the spillway gate hoist piers (and hoist bridge) can be allowed to fail. If the continuing operation of the gates depends on the continued viability of the hoist

regulating valves (especially low level release valves) to

maintain operating capability storage may have to be

quickly lowered if parts of the dam are damaged and need

remedial works or relief of hydrostatic loads.

bridge piers (and hoist bridge)

Spillway Gates

then carry out an appropriate

OBE: Gates to maintain normal operating capability.

dynamic analysis similar to that for the spillway piers. Combined static and dynamic loads may cause cracking but the piers must remain stable for overturning and sliding.

MDE: Gates retaining permanent storage at the time of an MDE should not fail to the extent where water from the storage is released in an uncontrolled manner. MDE should not cause the gates to distort to an

Note: the appropriate dynamic analysis required for spillway piers and spillway hoist piers may be part of an overall dynamic

extent that they cannot be opened or closed.

Spillway Piers

analysis for a concrete dam.

OBE: Carry out appropriate dynamic analysis for the spillway piers for earthquake loading in the

Access Bridge arid Piers OBE: Carry out appropriate dynamic » / of the pier and bridge ' The analysis may be

upstream/downstream and transverse directions.

Combined static and dynamic

88

ANCOLD Guidelines for Design

*4

Wke'

part of an overall dynamic analysis for an intake tower.

The method presented here is based on a

The pier and bridge system

domain using a suitable response spectrum. It uses an added mass representation of

should be examined for an

earthquake in the direction of the bridge access and perpendicular to the bridge access. Earthquake loads may be combined on a square root of the sum of the squares basis.

Combined static and dynamic

simplified dynamic analysis in the frequency

hydrodynamic effects due to surrounding (outside) water and contained (inside) water (in the case of wet towers). In addition, it includes the effects of tower/foundation interaction.

The steps of the method are:

loads should satisfy the appropriate Structures Code). MDE: If there are alternative means of access then the access bridge

can be allowed to fail. If there

(i) Select suitable response spectrum.

(ii) Compute the added hydrodynamic mass of water using Goyal and Chopra

(1989).

are no alternative means of

access then the bridge and piers should remain capable of carrying their design loads.

(in) Determine the structural properties of the tower

mass per unit height flexural stiffness

Access Roads

modal damping ratios.

OBE: Access roads to the dam and its appurtenant works should

remain passable immediately after the OBE. Therefore any likely land slip areas along the access roads should be checked

for stability. There should be no land slips either during or after

(iv) Compute natural periods and mode shapes for the first two modes of vibration.

(v) Determine the spectral accelerations for the first two modes of vibration from the response spectrum.

(vi) Compute the generalised mass and

an OBE.

generalised excitation terms using the MDE: Access roads to the dam and its appurtenant works may become impassable during or

immediately after an MDE. However, they should be easily

mass distribution and the mode shapes.

(vii) Compute the inertia forces using the

cleared. Therefore, while there

spectral accelerations, generalised mass and excitation terms, mass

may be land slips onto the

distribution and mode shapes.

roads, the roads themselves,

should not be allowed to collapse where there is no easy, alternative access route.

(viii) Add the inertia loads from the first two vibration modes on a square root of the sum of the squares basis.

(ix) Compute bending moments and shear forces from (viii).

8.3 Intake Towers 8.3.1 Analysis

(x) Design tower according to AS3600 The analysis method for intake towers is


described here in general terms only. A more

complete description of the method on which the following general description is based, is

8.3.2 Design Criteria

given by Chopra and Goyal (1991) and Goyal and Chopra (1989).

As discussed in Sub-section 8.2, an intake

tower is designed elastically for the OBE. Generally, it will not be necessary for the


89

tower to behave elastically during the MDE. In order to ensure that the tower can survive

intense ground shaking due to the MDE with limited damage, it should possess a ductility capacity greater than the ductility requirements imposed by the ground motion.

A suitable method for this is described in

Chopra and Liaw (1975) where a ductility factor of two is recommended (ratio of

maximum permissible displacement to the yield displacement).

/

90


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/


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USBR (1989(a)). Design Standard % Embankment Dams, No 13, Seismic Design and Analysis. USCOLD (1992). Observed performance of dams during earthquakes. US Corps of Engineers (1984). Rationalising the seismic coefficient method. Miscellaneous Paper GL84-13. US National Research Council (1985). Liquefaction of soils during earthquakes. National Academy Press, Washington DC. Waggoner, F., Plizzari, G. and Saouma, V.E. (1993). Centrifuge tests of concrete gravity dams. Dams Engineering, Vol IV, Issue 3, August. Westergaard, H.M. (1933). Water pressure on dams during earthquake. Transaction, ASCE, 98,

418-433 (see also Proc. ASCE, November 1931,1303-1318). White, W., Valliappan, S. and Lee, I.K. (1979). Finite element mesh constraints for wave propagation

problems. Proc. of the Third International Conference in Australia on Finite Element Methods. The University of New South Wales, Sydney, 531-539. William, K.J. and Wamke, E.P. (1975). Constitutive model for the triaxial behaviour of concrete. Int. Assoc. for Bridge and Struct. Eng. Procs., Vol 19.

Zangar, C.N. (1952). Hydrodynamic pressure on dams due to horizontal earthquake effects. US Bureau of Reclamation. Zienkiewicz, O.C., Clough, R.W. and Seed, H.B. (1986). Earthquake analysis procedures for dams -

state of the art. ICOLD/CIGB, Bulletin No 52. International Commission on Large Dams, Paris, 148pp. Zienkiewicz, O.C., Fejzo, R. and Bicanic, N. (1983). Experience in analysis plane concrete structures using a rate sensitive model with crack monitoring capabilities. Proc. of the Int. Conf. on Constitutive Laws for Engineering Materials, University of Arizona, Tucson, January. Zienkiewicz, O.C., Hinton, E., Bicanic, N. and Fejzo, R. (1980). Computational models for the transient dynamic analysis of concrete dams. Proc. Conf. on Design of Dams to Resist Earthquakes, Int. Civ. Eng., 171-178, London, October. Zienkiewicz, O.C., Paul, D.K. and Hinton, E. (1983). Cavitation in fluid-structure response (with particular reference to dams under earthquake loading). J.EarthquakeEng. and Struct. Dyn, 11.

Zienkiewicz, O.C. and Shiomi, T. (1984). Dynamic behaviour of saturated porous media: the generalised biot formulation and its numerical solution. Int. J.Num. and Anal. Meth. in Geomech, 8, 71-96.

Zienkiewicz, O.C., Valliappan, S. and King, LP. (1986). Stress analysis of rock as a non-tension material. Geotechnique 18, 55-56.

Zienkiewicz O.C. and Xie, Y.M. (1991). Analysis of the Lower San Fernando Dam failure under earthquake. Dam Engineering,2,302-322.


APPENDICES

/


APPENDIX A AUSTRALIAN NATIONAL COMMITTEE ON LARGE DAMS GUIDELINES FOR THE DESIGN OF DAMS FOR EARTHQUAKE TERMS OF REFERENCE OBJECTIVE To prepare guidelines for the design of dams in Australia for earthquake. The guidelines are to be sufficiently detailed to allow users to carry out preliminary assessments without reference to other material, and to give guidance on the methods available, their advantages and shortcomings for more detailed assessment where this is shown to be necessary by the preliminary assessment.

SCOPE OF GUIDELINES 1. Type of Dam - the guideline is to cover all dam types 2. Earthquake Loading in Australia • An overview of earthquake activity in Australia • Description of design earthquakes and acceptable damage design basis earthquake low probability events (eg. Maxm Credible Earthquake) concurrent load combinations relation to risk assessment dam and ancillary structures

• Calculation of design loadings from design earthquakes attenuation functions applicable in Australia effect of local geology, ground conditions, topography peak accelerations, spectra, components

• Methods for calculating design earthquakes probabilistic methods deterministic methods (relating to known faults etc) who should make this assessment

/


3.

Analysis of Effect of Earthquake Loading • For embankment dams

simplified "screening" methods to assess liquefaction potential and stability "second level" methods for analysing deformation "advanced level" analysis methods, eg. dynamic finite element data requirements for the above

acceptability criteria for the methods as related to stage of design, dam hazard rating and type • For concrete dams (gravity, arch, buttress etc)

simplified "screening" methods to assess stability "second level" methods for analysing stability, eg. 2D finite element "advanced level" analysis methods, eg. including crack propagation, 3D effects data requirements for the above

acceptability criteria for the methods as related to the stage of design, dam hazard rating and type • For tailings dams specific issues compared to embankment dams • For ancillary structures

simplified screening methods second level methods advanced level methods data requirements for the above

acceptability criteria as related to the stage of design, dam hazard rating and type of ancillary structure

4. Defensive Design for Earthquake Details of features which should be incorporated into dam designs which will reduce the probability of failure under earthquake. Issues include, for example, freeboard, filters, materials selection and zoning, drains.

5. Other Features of Earthquake Design Seiche - Landslides in reservoir

- Earthquakes induced by dam (not to be covered in detail)


APPENDIX B

- TYPICAL EASTERN AUSTRALIAN PEAK GROUND

ACCELERATION VS AEP - RESPONSE SPECTRUM FOR 1 IN 1000 AEP - MODIFIED MERCALLI SCALE


Earthquake Ground Motion Recurrence PEAK GROUND ACCELERATION RECURRENCE Tectonic Model using Esteva S Rosenblueth 1964 Attenuation

2 5 10 20 50 100 200 500 1000 2000 5000 10000 Return Period (Years)

Peak Ground Acceleration Recurrence

Earthquake Ground Motion Recurrence RESPONSE SPECTRUM - Trifunac 1980 Attenuation Tectonic Model Return Period 1000 years Probability of Exceedance = 0.50, Horizontal motion, No sediments

0.5 i 2 5

Frequency (hertz)

Response Recurrence for 1000 Year Return Period

TABLE H.4

MODIFIED MERCAULI SCALE, 1956 VERSION* Intensity Effocu

r*

v.t cm/t

M§ I. Not feit. Marginal and long-pertod effects of large earthquakes (for details see text). 3 II. Felt by persons at rest, on upper floors, or favorably placed.

0 0035-0 007

III. Felt indoors Hanging objects swing. Vibration like passing of light trucks. Duration estimated. May not be recognized as an earthquake.

0 007-0.015

IV. Hanging objects swing. Vibration like passing of heavy trucks; or

4 sensation of a jolt like a heavy ball striking the walls. Standing motor cars rock. Windows, dishes, doors rattle. Glasses clink. Crockery clashes In the upper range of IV wooden walls and frame creak.

V Felt outdoors; direction estimated. Sleepers wakened Liquids disturbed, some spilled. Small unstable objects displaced or upset. Doors swing,

1-3

0.015-0 035

3-7

0.035-0.07

7-20

0.07-0.15

20-60

0.15-0.35

close, open. Shutters, pictures move. Pendulum clocks stop, start, change rate.

VI. Felt by all. Many frightened and run outdoors. Persons walk unsteadily. 5 Windows, dishes, glassware broken. Knickknacks, books, etc., off shelves. Pictures off walls. Furniture moved or overturned. Weak plaster and masonry D cracked. Small bells ring (church, school). Trees, bushes

shaken (visibly, or heard to rustle—CFR). VII. Difficult to stand. Noticed by drivers of motor cars. Hanging objects quiver. Furniture broken. Damage to masonry D, including cracks. Weak chimneys broken at roof line. Fall of plaster, loose bricks, stones, tiles, cornices (also unbraced parapets and architectural ornaments—CFR). 6 Some cracks in masonry C. Waves on ponds; water turbid with mud.

Small slides and caving in along sand or gravel banks Large bells ring. Concrete irrigation ditches damaged. VIII. Steering of motor cars affected. Damage to masonry C; partial collapse. Some damage to masonry B; none to masonry A. Fall of stucco and some masonry walls. Twisting, fall of chimneys, factory stacks, monuments, towers, elevated tanks. Frame houses moved on foundations if not bolted down; loose panel walls thrown out. Decayed piling broken off. Branches broken from trees. Changes in flow or temperature of springs and wells. Cracks in wet ground and on steep slopes. IX. General panic. Masonry D destroyed; masonry C heavily damaged, sometimes with complete collapse; masonry B seriously damaged. 7 (General damage to foundations—CFR.) Frame structures, if not bolted, shifted off foundations Frames racked. Serious damage to reservoirs. Underground pipes broken. Conspicuous cracks in ground. In alluviated areas sand and mud ejected, earthquake fountains, sand craters. X. Most masonry and frame structures destroyed with their foundations. Some well-built wooden structures and bridges destroyed. Serious 8 damage to dams, dikes, embankments. Large landslides. Water thrown on banks of canals, rivers, lakes, etc. Sand and mud shifted horizontally on

60-200

0 35-0.7

200-500

0.7-1.2

beaches and flat land. Rails bent slightly. >1 2

XI. Rails bent greatly Underground pipelines completely out of service. XII. Damage nearly total. Large rock masses displaced. Lines of sight and level distorted Objects thrown into the air.

From Fig. 11.14

NOTE. Masonry A, B, C, D. To avoid ambiguity of language, the quality of masonry, brick or otherwise, is specified by the following lettering (which has no connection with the conventional Class A. 8, C construction) ¦ Masonry A. Good workmanship, mortar, and design, reinforced, especially laterally, and bound together by using steel, concrete, etc.; designed to resist lateral forces. ¦ Masonry 8* Good workmanship and mortar; reinforced, but not designed to resist lateral forces ¦ Masonry C. Ordinary workmanship and mortar, no extreme weaknesses such as non-lied-m corners, but masonry is neither reinforced nor designed against horizontal forces. ¦ Masonry 0* Weak materials, such as adobe, poor mortar, low standards of workmanship, weak horizontally •From Rlchter|l958|1 Adapted with permission of W H Freeman and Company Âverage pt/ak ground velocify, cm/s f Average peak acceleration (away from source) ^Magnitude correlation

APPENDIX C

EXTRACTS FROM CANADIAN DAM SAFETY GUIDELINES


Dam Safety Guidelines

CDSA

1.4 CLASSIFICATION OF DAMS Requirement: Each dam shall be classified In terms of the reasonably foreseeable consequences of failure. Each water retaining structure, including water passages, shall be classified separately. Each dam should be classified in accordance with the consequences of failure. The

classification constitutes the basis for analysing the dam's safety and setting appropriate levels of surveillance activities. Table 1-1 represents a commonly-accepted classification

system which is based on the potential loss of life and economic damages associated vrfth dam failure. This classification system is used to link the consequences of failure to the requirements contained in Sections 2 through 10.

Aftemative classification systems may be adopted for interpreting and addressing the requirements for dam surveillance and dam safety reviews, as set out in Sections 2 and

3 of these Guidelines. Such classification systems may incorporate the physical characteristics of the dam, its condition and the perceived risk of its failure, as well as the consequences of failure. Appurtenances may be classified and evaluated separately. Thus the water passages could be in a different category from the dam, depending on the consequences of failure.

If warning systems are considered to reduce the potential loss of life, the reliability of such warning systems must be incorporated into all analyses and evaluations. The consequence categories listed in Table 1-1 are based on the incremental losses which a failure might inflict on downstream or upstream areas or at the dam. "Incremental losses" are those over and above losses which might have occurred for the same natural event or conditions, had the dam not failed. The distinction between consequence categories, and the link with safety requirements, is intended to reflect society's values and priorities in allocating and distributing resources and funds to be used for protecting and saving lives, and for safeguarding property. The incremental consequences of dam failure should be evaluated in terms of:

Loss of life Economic value of other losses and/or damage to property, fadi'rties, other utilities and dam, as well as loss of power generation or water supply. Where appropriate, costs are assigned to social, cultural and environmental impacts. Other less quantifiable consequences related to social, cultural and environmental damages.

The most severe consequences should prevail - if economic losses are Very High and loss of life is High, the dam would be classified as a Very High Consequence dam. / Evaluation of potential losses, both with and without dam failure, should be based on inundation studies and should consider existing and anticipated future downstream development and land uses. At the same time, the appropriate study level of inundation would depend on the potential consequences of failure. For dams where there is uncertainty about the consequences of a dam break, a simplified and conservative

January 1, 1995

Page 1-3

CDSA

Dam Safety Guidelines mmmmmmmmvinimw iMMVk* *

analysis should be used to make a preliminary assessment. If this analysis demonstrates a potential hazard, a more sophisticated analysis should then be undertaken. In the case of dams where the consequences of failure clearly fall within the "Very Low" category, a formal inundation study is not required. A dam may be in one category for flood hazard and a different category for earthquakes, depending on the incremental damage attributable to dam failure from each cause. A screening level estimation of the incremental consequences of failure may be appropriate for a dam to be classified in the Low Consequence category. However, if

a dam is likely to be classified in the High or Very High Consequence categories, the evaluation of incremental consequences of failure should be based on she-specific analysis, and may require detailed site investigation. Consequences of dam failure due to earthquakes should be based on average discharge conditions and maximum normal operating levels. Consequences attributable to reservoir slope failure or slope-failure-induced waves should be based on average discharge and maximum normal operating levels, unless the slide would have been induced by extreme rainfall associated with an extreme flood.

January 1, 1995

Page 1-4

CDSA


TABLE 1-1

CONSEQUENCE CLASSIFICATION OF DAMS

CONSEQUENCE CATEGORY

POTENTIAL INCREMENTAL CONSEQUENCES OF FAILUREl*1 LOSS OF LIFE

ECONOMIC, SOCIAL, ENVIRONMENTAL

VERY HIGH

Large increase expected ^

Excessive increase in sodal, economic and/or environmental losses.

HIGH

Some increase expected ^

Substantial increase in sodal, economic and/or environmental losses.

LOW

No increase expected

Low sodal, economic and/or environmental losses.

VERY LOW

No increase

Small dams with mtnimal sodal, economic and/br environmental losses. Losses

generally limited to the owner's property; damages to other property are acceptable tosodety.

[a] Incremental to the impacts which would occur under the same natural conditions (flood, earthquake or other event) but without failure of the dam. The type of consequence (e.g. loss of life, or economic losses) with the highest rating determines which category is assigned to the structure.

[b] The loss-of-life criteria which separate the High and Very High categories may be based on risks which are acceptable or tolerable to society, taken to be 0.001 lives per year for each dam. Consistent with this tolerable societal risk, the minimum

criteria for a Very High Consequence dam (PMF and MCE) should result in an annual probability of failure less than 1/100,000.

January 1, 1995

Page 1-5


CDSA

MMNMMMMMMMimMliilMlttWrtMiliMMff*-

1.5 SELECTION OF SAFETY CRtTERIA Requirements: The dam, alongj with Its foundation and abutments, shall have

adequate stability to safely withstand extreme loads as well as the normal design loads.

The selection of loading criteria for extreme loads shall be based on the consequences of failure of the dam. Methods to determine appropriate normal design loadings and factors of safety are covered in Sections 5 through 9 of this document. Sections 5 and 6 address earthquake loadings and floods, respectively. To select criteria for extreme events, a risk-based approach may be used. The principle is that a dam whose failure would cause excessive damage or the loss of many lives should be designed to a proportionately higher standard than a dam whose failure would result in less damage or fewer lives lost. In assessing the safety of an existing dam,

probabilistic risk analysis (PRA) methods can help verify that qualitative factors such as internal erosion and spillway debris blockage are not overlooked and that they receive

attention commensurate with their contribution to the failure probability. The level of safety of a dam can sometimes be improved by addressing conditions less severe but more likely than those associated with such extreme events as the Maximum Credible

Earthquake (MCE) and the Probable Maximum Flood (PMF). Criteria for extreme events other than floods and earthquakes should be consistent with the levels required for flood and earthquakes.

January 1, 1B95

Page 1-6

CDSA Dam Safety Guidelines

5.0 EARTHQUAKES M The use of criteria other than those indicated in this document may be appropriate or necessary, taking into account specific conditions arising at some dam projects, and to permit development in the application and use of new knowledge and improved techniques.

Requirement: Dams shall be designed and evaluated to withstand ground motions associated with a Maximum Design Earthquake (MDE), without release of the reservoir.

Selection of the MDE for a dam shall be based on the consequences of dam failure. The MDE is usually represented by the most severe ground motion which has been selected for design or safety evaluation of the dam. Site-specific ground motion parameters required for design or evaluation are determined from the MDE. For a given site, the MDE should increase with increasing consequences of dam failure,

as illustrated in Table 5-1. For a given Annual Exceedance Probability (AEP), the MDE from site to site may also vary with the tectonic setting of the site and the distance from the earthquake source. In some cases, the MDE selection may be based on seismic loading that could be triggered by human activity, some examples being extraction or injection for oil fields, and reservoir-induced seismicity. The development of site-specific seismic parameters such as ground velocities, accelerations and response spectra, shall be derived from the criteria for design earthquakes in Table 5-1. Methods to achieve this should conform to currently accepted practice in different regions of Canada. Derivation of seismic parameters should be undertaken or supervised by persons with appropriate specialties in earthquake engineering. Well built embankment dams that are sited on firm non-fiquefiable foundations and that do not incorporate large bodies of materials which, if saturated, might lose most of their strength during earthquakes, can be designed and evaluated using seismic coefficient methods (pseudostatic analysis) under the conditions outlined in Section 8.1. The seismic coefficient should reflect the seismicity of the dam site and it can be obtained from zoning maps created for that purpose.

/

[a] This section addresses criteria for design earthquakes only. The requirements for structural resistance to the earthquake are presented in Sections 8 and 9.

January

1.

1995 Page

5-1


CDSA

mmmommmmmmmimiitmimmsm

TABLE 5-1

USUAL MINIMUM CRITERIA FOR DESIGN EARTHQUAKES

CONSEQUENCE CATEGORY

MAXIMUM DESIGN EARTHQUAKE (MDE) DETERMINISTICALLY DERIVED

PROBABILISTICALLY DERIVED (Annual exceedance probability)

Very High

High

Low

MCE MM [d]

50% to 100% MCE[e]

. [g]

1/10,000 M M

1/1000 to 1/10,000 W

1/100 to 1/1000 k!

[a] See Section 1.4 for consequence classification. [b] For a recognized fault or geographically defined tectonic province, the Maximum Credible Earthquake (MCE) is the largest reasonably conceivable earthquake that appears possible. For a dam site, MCE ground motions are the most severe ground motions capable of being produced at the site under the presently known or interpreted tectonic framework. [c] In Hydro-Quebec's practice, the MDE for Very High Consequence structures involves a combination of deterministic and probabilistic approaches that reflect current knowledge of seismo-tectonic conditions in Eastern Canada. Hydro-Qu6bec's deterministically derived MDE magnitude is the maximum historically recorded earthquake, increased by one-half magnitude, while their probabilistically derived earthquake has an estimated probability of exceedance of 1/2000. [d] An appropriate level of conservatism shall be applied to the factor of safety calculated from these loads, to reduce the risks of dam failure to tolerable values. Thus, the probability of dam failure could be much lower than the probability of extreme event loading. [e] MDE firm ground accelerations and velocities can be taken as 50% to 100% of MCE values. For design purposes the magnitude should remain the same as the MCE [f] In the High Consequence category, the MDE is based on the consequences of failure. For example, if one incremental fatality would result from failure, an AEP of 1/1000 could be acceptable, but for consequences approaching those of a Very High Consequence dam, design earthquakes approaching the MCE would be required. [g] If a Low Consequence structure cannot withstand the minimum criteria, the level of upgrading may be determined by economic risk analysis, with consideratten of environmental and social impacts.

mmMXfwtttMttrfi'itftrtmrfm pti n lyrrfr

January 1, 1995

Page 5-2

APPENDIX D

ADDITIONAL INFORMATION ON ACCEPTABLE RISKS


REFERENCES 1. ANCOLD (Australian National Committee on Large Dams), "Guidelines on Risk Assessment", January 1994. 2. Appleyard, L.D. and Narwar, G., "Risk Assessment in the Development of

Serviceability Based Design Criteria for Small Buildings", Proceedings of the Conference on Probabilistic Risk and Hazard Assessment, Newcastle NSW, 22-23 September 1993. 3. BC Hydro, "Guidelines for Consequence Based Dam Safety Evaluations and Improvements", Report No. H2528, Hydroelectric Engineering Division, Bumaby, BC, Canada, August 1993. 4. CERIA (Construction Industry Research and Information Association), "Rationalisation of Safety and Serviceability Factors in Structural Codes", Report 63, London, July 1977.

5. DOP (Department of Planning), New South Wales, "Risk Criteria for Land Use Safety Planning", Hazardous Industry Planning Advisory Paper No. 4, 1992. 6. Higson, D.J., "Nuclear Safety Assessment Criteria" Nuclear Safety, Vol. 31, No. 2,

April-June 1990. 7. HSE (Health and Safety Executive, United Kingdom), "Risk Criteria for Land Use Planning in the Vicinity of Major Industrial Hazards", London, Her Majesty's Stationery Office, 1989. 8. HSE (Health and Safety Executive, United Kingdom), "Quantified Risk Assessment: Its Input to Decision Making", London, Her Majesty's Stationery Office, 1989. 9. Kletz, T.A., "The Application of Hazard Analysis to Risks to the Public at Large". Paper presented to World Congress of Chemical Engineering, Amsterdam, 1976. 10. Reid, S.G., "Practical Procedures for Setting Standards", Lecture 8, One Day Post¬ graduate Course on Engineering Risk Assessment, University of Sydney, 8th March,

1991. 11. Technical Advisory Committee on Water Defences, Centre for Civil Engineering Research and Codes, "Probabilistic Design of Flood Defences", Gouda, The Netherlands, June 1990. 12. The Royal Society of London, "Risk Assessment: A Study Group Report", The Royal Society, London, 1983.

13. Whitman, R.V., "Evaluating CaJculated Risk in Geotechnical Engineering", Journal of Geotechnical Engineering, ASCE, Vol. 110, No. 2, February 1984, pp. 145-188. 14. Wilson, R. and Crouch, E.A.C., "Risk Assessment and Comparisons: An Introduction", Science, Vol. 236, April 1987.

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potential number of premature fatalities in the event of a dam incident

/

COMPARISON OF PROPOSED INDIVIDUAL AND SOCIETAL RISK CRITERIA AND RISK CRITERIA USED IN THE NETHERLANDS AND UNITED KINGDOM (BC HYDRO, 1993, AND ANCOLD, 1994).

COMPARATIVE RISK LEVELS The risk levels quoted in the following Individual Risk table are the additional increment of risk arising from the activity in question; except for the first two risks. The risk levels are as quoted in the references given in brackets (see Attachment C for references).

Individual Risk is the total additional increment of risk imposed by a facility (such as a dam) on a particular person. All such risks may be averaged over a particular group.

Societal Risk is a measure of society's aversion to loss of life, especially to catastrophes involving large loss of life. It has no regard to the identities of persons and is related to the occurrence of an event at a facility.

COMPARATIVE INDIVIDUAL RISK LEVELS INDIVIDUAL RISK SITUATION Male, age 50, UK (7) Female, age 20, UK (7)-

Rock climbing, UK (7) Smoking 20 cigarettes per day (6) Smoking 20 cigarettes per day, persons at risk (5) Commercial Diving (3)

Deep Sea Fishing, US (10) Parachuting, US (10) Offshore Oil/Gas Workers (12)

Hang GUding, UK (7) Air Crew, persons at risk (4) Mountaineering, persons at risk (14) Drinking, alcohol, persons at risk (5) Car Travel, persons at risk (2) Road Accidents, US (10) Quarry Workers (10) Car Travel, 10,000km/year, British Columbia (3)

Coal Mining, US (10) Road Accidents, whole population (6) Air Pollution, Eastern USA (14) Coal Mining, UK (7) /

chance per person per year

1 in 1 in 1 in 1 in 1 in 1 in 1 in 1 in I in 1 in 1 in 1 in

1 1 1 1 1 1 1 1 1

in in in in in in in in in

55 3000 125 200 200 350

360 530 600 670 1,000 1,660 2,600 2,800 3,300 3,400 3,500 4,800 5,000 5,000 9.500

Commercial Air Travel, whole population (2)

1 in

Road Accidents, UK (7)

I in

Construction Worker, UK (7)

I in 1 in

Drinking I bottle wine/day (9) Swimrrung, persons at risk (6) Pedestrians struck by Car, whole population (5)

Air Travel (3) Train Travel, persons at risk (6)

1 in

1 in 1 in 1 in

Lightning strike, UK (7)

1 in I in I in 1 in 1 in 1 in 1 in 1 in I in I in 1 in I in 1 in 1 in 1 in I in 1 in 1 in 1 in 1 in 1 in 1 in 1 in 1 in 1 in

Lightning strike, whole population (6) Lightning, whole population (5)

1 in 1 in

Petrochemical, criterion, average to the public (9)

1 1 1 1

Drowning, US (10) Homicide, whole population (5) Domestic electrocution, Canada (3)

Air Travel, persons at risk (6) Petrochemical, criterion, maximum to public (9)

Dam Failure, limit criterion, exposed persons (I) Industrial plants, HSE limit, most exposed group (7) Nuclear plants, ICRP limit, to public (6) Air Travel, US (10) Drowning, UK (7) Electrocution (14) Building fires, Australia (10) Railway travel, US (10) Electrocution, whole population (6) Earthquake, California (9) Dam Failure, objective criterion, exposed persons (1) Industrial plants, DOP objective criterion (5) Industrial plants, HSE objective, most exposed roup (7) Argentine nuclear standards, limit criterion(6)

Nuclear plants, USNRC objective, prompt fatality (6) Nuclear reactor accidents (3)

Building collapse, persons at risk (2) Structural collapse, UK (10)

Nuclear power station at 1km, UK (9) Nuclear power station at boundary, US (9) Flooding, North Sea Dikes (9) Building Collapse, whole population (2) Structural Failure (4) J Meteorite strike, whole population (6) f Meteorite strike, whole population (5)

in in in in

1 in

1 in I in 1 in

10,000 10,000 10,800 13,300 20,000 28,600 30,000 33,000 33,300 50,000 65,000

100,000 100,000 100,000 100,000 100,000 110,000 170,000 190,000 250,000 250,000 330,000 590,000 1,000,000 1,000,000 1,000,000 1,000,000 2,000,000 2,500,000 to 5,000,000 5,000,000 7,000,000 10,000,000 10,000,000 10,000,000 10,000,000 10,000,000 10,000,000 10,000,000 10,000,000 10,000,000 1,000,000,000 1,000,000,000

/ Abbreviations:

DOP HSE ICRP USNRC

Department of Planning NSW Health and Safety Executive, UK International Commission for Radiological Protection US Nuclear Regulatory Commission.

COMPARATIVE SOCIETAL RISK LEVELS L - Limit O - Objective NA - Not applicable SOCIETAL RISK CRITERIA - chance per facility per year

SOURCE 10 Lives

1,000 Lives

100 Lives

ANCOLD for dams (1)

1 . 10,000(L) 1 • 250,000(0)

BC Hydro for dams (3)

1 ' 10,000(L)

1 . 100,000(L)

I : 1,000,000(L)

No objective set

No objective set

No objective set

UK Nuclear

No Limit given

No Limit given

No Limit given

Industry (5)

1 : 200,000(0)

1 : 10,000,000(0)

1 : 100,000,000(0)

Netherlands Planning

1 : 100,000(L) I : 10,000,000(0)

1 : 10,Q00,000(L) 1 : 100,000,000(0)

1 • 100,000,000(L) 1 : 1,000,000,000(0)

NA NA

NA

Surge Barrier (11)

NA NA

Environmental Norms

1 : 10,000(L)

I : 100,000,000(L)

Province of Groningen (11)

1 : 100,000,000(0)

1 : 1,000,000(L) 1 : 1,000,000,000(0)

Netherlands Liquefied Petroleum Gas Integral

1 : 100,000(L) 1 : 10,000,000(0)

1 : 10,000,000(L) 1 : 1,000,000,000(0)

Too small

1 : 3,000,000(0)

1 :7,000,000(0)

1 : 12,500,000(0)

Rules (5) Eastern Scheldt Storm

1 • 140,000(L) 1 : 12,500,000(0)

1 : 10,000,000(L) 1 : 1,000,000,000(0)

I : 10,000,000 (O)

Too small

1 : 1,000,000,000(L)

Note (11) Sizewell B Nuclear Station,

UK (8) Thames Barrier (8) Rail Tunnel, England to France (3) Canvey Oil Refinery, Chemical Plant, UK Second Report (8)

1 : 1,000(0)

1 :100(0)

1 : 1,100(0)

1 : 1,250(L)

1 :2,000(L)

1 : 5,000(L)

NA

NA

NA

1 : 100,000 (actual)

1 : 10,000,000 (actual)

1 : 1,000,000,000 (actual)

1 :70,000,000 (actual)

1 : 1,000,000,000 (actual)

Too small

US (6) Dajn Failure US (13)

1 : 10,000

1 :25,000 (estimated)

1 : 100,000

Sequoyah Nuclear Plant,

US (6) Grand Gulf Nuclear Plant,

(estimated) Notes:

(estimated)

Risks lower than 1 : 100,000,000 have been rounded up to the nearest whole order. Canvey Complex, UK is an existing facility, important to the national economy. The accepted risks are regarded as high.

- - >'


Dam engineers in Australia have been conscious of earthquakes for many years, but it was the earthquakes at Tennant Creek in 1988, which were M6.3,' M6.4 and M6.7, with a total fault scarp length of 32 km which raised the question most acutely as to whether dams in Australia could be subject to large earthquakes, and if so, could they withstand them without resultant loss of the facilities and lives, property, and environmental values downstream. In recognition of the need to provide some guidance to dam engineers and owners in Australia, the Australian National Committee on Large Dams

(ANCOLD) established a Working Group to prepare Guidelines for Design of Dams for Earthquake. These guidelines cover all types of dams, including tailings dams, and apply to existing and new dams. They cover the selection of the design earthquake, analysis and design of embankment and concrete dams, and appurtenant structures. Whilst specific to Australian considerations, the majority of this guideline could be applied to dam structures throughout the world.


ij

Guidelines for Design of Dams for Earthquake

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