LOADS AND PERFORMANCE DESIGNING
FOR
23
LOADS
Present design practice for transmission ,structure foundations relies upon the use of methods and formul_as which attempt to define the ultimate or failure capacity of the foundation in the various soil or rock types encountered. Foundations are also designed to a deflection criteria, usually at a working load. Soil parameters are determined from past experience in the area (if any) coupled with various amounts of geotechnical study and field exploration. The soil values used for design are generally conservative depending upon the degree of actual data and testing that went into their determination. The applied loads for the foundations generally include a small additional overload factor applied to the structure ultimate loads except for NESC loads for which the specified NESC overload factors for foundations are used. The result is a foundation which the engineer believes will sustain the applied factored loads whenever they occur. The use of a load and resistance factor design (LRFD) format as presented in the ASCE Transmission Line Structural Loading Guide will allow varying both the load factor and the foundation strength factor to suit the given conditions. Load factors (one or larger) are applied to account for the statistical nature of variation of the climatic loads as well as to provide extra reliability in important lines or greater safety for conditions where failures can injure workmen. Load factors can also be used to control a sequence of failure, thus foundation can be designed the structure it supports.
by the use to withstand
of load greater
factors a loads than
Resistance factors (one or smaller) are applied to the strength of the component and generally reflect the variability of the strength and the confidence in the knowledge of the material properties or the accuracy of the design methods. Thus a foundation design strength value should normally have a smaller resistance factor than would be assigned to a more uniform component such as steel. The advantage of means to design for also to identify which In order to achieve any transmission line need to be determined. equations proposed de fIe c t ion sh 0 u I d be result is an average value. Additionally,
the LRFD method is that it provides a a desired probability of failure and component is the more likely to fail. this for the design of foundations or component, strength resistance factors For foundation design, methods or for determining the strength or s p e c i f i cas tow he the r the de term in e d strength value or a minimum strength sufficient tests should be performed
24
TRANSMISSION LINE TOWERS FOUNDA TrONS
to eventually establish a data base from which the degree of variability of the results or a coefficient of variation can be established. Knowing the coefficient of variation allows the engineer to select the strength resistance factor which provides the degree of reliability which is desired. SUMMARY Transmission other structures different safety reflected in the
line structures are unique compared to such as bridges or buildings. They require and reliability criteria which should be design of the foundations.
The reliability based LRFD method can be used to assign different reliabilities to the foundations and other transmission line components and provides a means to account for the degree of variation of actual foundation strength versus the calculated strength. ACKNOWLEDGEMENT Much Sections
of 1
the and
material 2 of
in the
this IEEE
paper is Trial-Use
based Guide
upon for
Transmission Structure Foundation Design which was prepared by a joint ASCE/IEEE committee and is currently under revision by a joint ASCE/IEEE committee which includes the authors of this paper.
APPENDIX
1.
- REFERENCES
ASCE Foundations Subcommittee/IEEE of Foundations for Transmission Trail-Use Guide for Transmission Design, Institute of New York, New York.
2.
Electrical
Subgroup Structures, Structure Electronic
Committee on Electrical Transmission Transmission Line Structural Loading Society of Civil Engineers, New York,
on Design 1985, IEEE Foundation Engineers,
Structures, 1984, Guide, American New York.
CONSTRUCTION AND DESIGN OF FOUNDATIONS FOR FRENCH OVERHEAD POWER TRANSMISSION NETWORK
* M. GAGNEUX
** J.L. LAPEYRE
This paper summarizes present conception and design of foundations for90 to 400 kV overhead lines in France. Firstly principles of geotechnical studies in use are presented : various types and number of soil investigations are discussed. Then several kinds of foundations employed are described : - pad and chimney foundations for towers in soft soils ; block foundation in case of rocky soils - present tendency in use of driven piers for separate foundations or towers ; - roundations for single poles. At last, designs presented.
and
safety
coefficients
of
such
footing
foundations
are
O. Introduction In 1985 Electricite de France commissioned (in tower files) : 950 km of 400 kV overhead lines, 200 km of 220 kV overhead lines, and 450 km of 63/90 overhead kV lines. The coming years will witness principally an increase in the construction of 63/90 kV lines. - The foundation average cost represent 7 to 15 % of the construction costs. Since 1970 design and technological conceptions or foundations have considerably progressed. This paper presents the major developments in these diverse fields. 1. Soil reconnaissance Soil investigations are performed essentially for high voltage (63 or 90 kV) or extra-high voltage (225 or 400 kV) structure. These structures are chiefly four-legged lattice towers which apply tension/compression loads to the soil, and since about ten years, single member supports (called "Muguet" in France) of aesthetic finality but of rather limited use. 11. Originally: the "penevane" It will be seen subsequently that the calculation method, developed in the 1970's and used in France, requires the knowledge of limit characteristics of the soil : C, ~ and Being conscious or precautions to be taken for determination of C and ~
I.
ELECTRICITE DE FRANCE : * Engineer. Service du Transport - Centre d'Equipement du Reseau de Transport 92068 PARIS LA DEFENSE CEDEX 48 - FRANCE. ** Engineer. Direction des Etudes et Recherches 1 avenue du General de Gaulle 92141 CLA}~RT CEDEX - FRANCE.
25
26
TRANSMISSION LINE TOWERS FOUNDA nONS
in laboratory, emphasis was put on the development of an "in-situ" measuring method that would be convenient, rapid and if possible unexpensive. It is under such circumstances that the "penevane" whose design and prototype are due to Mr TRAN VO NIEHM L8] , was defined. This apparatus, that combines a dynamic penetrometer and a scissometer, was being used during few years in the early 1970's by contractors entrusted with line construction. Comparative tests had enabled correlations to be established between the cohesion, angle of internal friction and the dynamic and scissometric characteristics provided by the penevane .. However, this apparatus was not able to keep all its promises mainly because of insufficient penetration power. Developing only an energy of 10 daN.m, its driving-in was often impeded not only by compact layers even of small thickness, but also by the presence of small-sized solitary boulders. Facing such a state of things, decision was then taken to come back to C and cp measurements, using geotechnicians for these measurements. 12. Officialization
and codification
of soil studies
Calling almost systematically uppon the geotechnical engineer prior to any choice of foundations, occurred only by the late 1970's. If at the outset the idea was to preemptively obtain the values of cohesions and angles of internal friction needed for the design of shallow foundations, it appeared very soon that the soil studies could enable us to obtain a large set of informations both qualitative and quantitative. They may for instance provide valuable indications on problems associated with the actual execution of excavations (tools to be employed) and on the time stability of their walls. Besides, with the growing number of pile supported towers, it is indispensable that quantitative data be available, for their sizing and also for the choice of the pile type to be made use of. Starting from the simple idea that each line forms a unique structure, we try both to limit the number of measurements and soundings, in order to obtain not too significant costs, and to carry out a sufficient number of measurements so that the results thus obtained should not be contested. 13. Qualitative study The qualitative study prepares the measurement campaign which will end with the design or foundations. It is based on geological maps, information gathered by geotechnicians during earlier soundings and on a obligatory field reconnaissance campaign (auger sampling, shovelled pits, dynamic penetrometer, seismic-refraction). It makes it possible, first of all, to get an insight into the nature of ground layers encountered all along the line and hence to attract attention to the zones where shallow foundations can be implemented as well as to those where their installation is to be procluded. Next, it leads to define homogeneity zones, namely sections or groups of sections where the character of the soil is markedly the same for all supports that are to be installed. It endeavours also to gather informations relevant, for example, to hydrogeology, as well as to the difficulties of actual execution of excavations as to the kind and number of measurements to be considered
in the quantitative
stage.
OVERHEAD TRANSMISSION NETWORK
27
14. Quantitative study The quantitative study consists essentially in taking samples on selected sites to measure in laboratory the C and values and to perform pressuremetric measurements with a view to determine the permissible compressive stresses and as well as the characteristics being of use for the calculation of special pile foundations whether acted on by tensile/compressive or overturning stresses.
¢
We attract attention to the fact that the importance of cohesion in the calculation of the block has led us, under the present conditions, to consider, in the design, the long term (or drained) characteristics C', (usely, C' is notably less than Cu undrained cohesion) and this because durations of excavations opening in frequently encountered, silty soils are rapidly leading to a perceptible reduction in the cohesion of these grounds.
¢'
At the present time, generally one support location every 4 to 10 is concerned either by a non-destructive sampling or by a pressuremetric sounding to which are being added the elements contributed by the qualitative study, in particular by dynamic penetrometers. Between a sounding at each foot tower as some would desire - (an ideal but too expensive solution) - and the present practice, there exists a range where the extrapolation of obtained results to unsounded locations should be done. It is why we recommend that prior to the beginning of works, a meeting systematically takes place between all the actors that contribute to the choice or actual execution of foundations : prime contractors, geotechnicians, purchaser. The purpose of this meeting, after a last examination of adopted solutions, is for one thing to advocate under which conditions and on which assumptions the choice of foundations was made and more particularly at locations where no soil-investigation was performed. For another, it is important to examine for these locations, the elements that may lead to question again some hypothesis adopted at the outset : thickness of different nature of the surface layer, presence or not of water and to derive therefrom eventual modifications of the original foundations. In spite of this proceeding by successive approximations it may happen, nevertheless, that the contractor is in presence of a situation which j under these was not considered during previous proceedings circumstances, the geotechnician is requested, to carry out a complementary study with a view to define very rapidly the changes to introduce in the initial project. 2. Diverse kinds of foundations 21. Foundations for four-legged lattice towers 211. Shallow foundations (pad and chimney foundations These still represent at the present time most foundations used for 63/90 kV, 225 kV, and 400 kV lines. An example for medium soil is given in Figure 1 for an uplift force of 700 kN. These pad and chimney foundations are still widely used, because they are economical and can be carried out
28
TRANSMISSION
LINE TOWERS
FOUNDATIONS
without particular implementation means. For lines based on this type of blocks, the foundation cost represents, an average of 10 to 12 % of total line cost. 212. Deep foundations Uplift capacities are to day greater than in the past, and a significant advance in this foundations design can be nowadays noted. The cost of this kind of foundation represents around 15 to 17 % of the overall line cost. In the early days of EDF's existence, special foundations like driven-piles were only intended for poor-grade soils in which all shallow foundations solution was to be precluded, things have very noticeably changed since. Pile or group of piles appear now more and more often as the rival of the shallow foundations, because of the increase in the uplift stresses. The drilled and cast piers used until the early 1950' s were rapidly superseded by the HULLER pile introduced in France. This pile with metallic core is driven and is surrounded by exterior mortar jetting (Figure 2). The development of HuLLER system, especially since the 1970's led to the implementation or piles with growing transverse sizes. Starting from metallic cores of diameter varying from 250mm to 450mm and requiring, in order to ensure the junction with the support, the construction of a reinforced concrete bonding block always expensive, the special roundations companies are now implementing piles or diameters larger than 1000mm. The significant sizes, in addition to the mechanical advantage offered by a great top inertia to absorbe high secondary moments, permit furthermore direct interconnection of the tower base angle inside the tube. Another major advantage lies in the fact that only one of these injected piles allows tensile or compressive stresses more than 4000 kN. Driven piles, whether they have small or large transverse sizes, can be implemented in all the soils ranging from low characteristics to rairly compact ones (pressuremeter limite pressure lower than 2 to 2.5 ~Wa). In case of strongly consolidated grounds such as rocks, the possibility of achieving drilled and injected micropiles (100 to 4S0mm in diameter) permits realization of foundation without impairing the natural environment by the use of explosives. Thus, regarding the pile, there always exiqts a solution in this technical field that can be implemented whatever the caracteristics of the soil.
22. Foundations for single poles The shallow foundations, for reinforced concrete poles or metallic poles (for 63/90 kV voltages) are still being widely used in the
OVERHEAD
TRANSMISSION
NETWORK
01
29
I [0,10
I10.20
I I
I
j
2
0.15
, x
o
I
I
,
j c
~f, i
.c:1I r
a
Figure 1
Shallow foundation block with base plate : 700 kN.
uplift force
0.7 D : 2.90 m a 1.85 m
CP1 CP?
c = 1.55 m x=0.15m h = 0.5 m
h1 = 0.4 m
m
1.10 m
h3
1. 10
=
m
construction of such lines. However, the implementation of single pole supports for 225 and 400 kV voltages has led to a change in the design of these blocks through the use of deep foundations. At the outset, for these 225 and 400 kV structures, the foundations group of injected MULLER type piers, as well as the bond with the tower - (flange and rods embedded in a reinforced concrete block) - formed complicated and expensive assemblies. Very soon simplifications were applied : the groups of piers were replaced by a driven cylindrical metallic tube of large diameter ( > 1200mm), the bond between the tower and this foundation being provided by a flange welded onto a tubular cap covering the upper part of the foundation tube over a height of about 2.50m (Figure 3). > 1800rnrn) allowed The increase in diameter of driven shells afterwards, in many cases, the base section of the tower to be directly sealed within its foundation (Figure 4). This bind of tower-foundation interconnection is also employed for the 63/90 kV lines. In spite of these improvements, the cost of such constructions can still be from 15 to 30 % of the line cost.
¢
(9
3. Calculation methods 31. Calculation of foundations for four-legged 311. Shallow foundation blocks
towers:
Researches were conducted as early as 1963, by EDF in close cooperation with the Grenoble University. It is not our purpose to deal in detail with the calculation method proceeding from these studies, called "c, method" a description of which can be found in [lJ, [6J, [7J.
¢'
I
30
Coupling conneciion
¢ 2.50
0 ~OOOmm
In\eciion
Injected driven pier of 250 to 1000 mID diameter
31
OVERHEAD TRANSMISSION NETWORK Maximum
I
applied overturning moment: 80CXJmkN
i1T I I I I I
Si", m,m'" '",PO"
Epoxy resin injection
~.~
..,..,...'l':'1K.,;...,
aa r<"I
Bond resin
I
~I C\J
Soil in piece
Materiel steel E36
l I
Shell (01800 mm thickness
20mm
20
Figure 3 : Driven pier with cap (Epoxy resin seal) Let us merely indicate that these various researches, completed by full-scale tests (carried out in several countries), enabled to show that during the uplift tests on foundations, surface sliding within the body of soil appears ; these surfaces whose intersection with vertical planes (slip lines) was assimilated to straight lines sloped away from the vertical by an angle a (Figure 5). The wedge of soil inscribed inside the faces BC and AD remains linked with the foundation during its displacement. It is then possible to consider that the faces BC and AD are under passive soil pressure. The determination of therein developing stresses is performed by taking into account the theories of plasticity and limit equilibrium, applied to soils. Furthermore, it should be pointed out that these researches gave rise to an important concept : that one of critical depth D beyond which If b (see the shearing lines no longer propagate to the ground surface. Figure 5) is the side of a square base foundation (or the small side of a rec tangular base) this value D is such that : D /b ~ 2.5. The formula which gives the ultimate up~ift resistance of aCfoundation and which we use today for determining our shallow foundation blocks is
32
TRANSMISSION
LINE TOWERS
FOUNDATIONS
Sealing
concrete
Inspection
aperture
'" .!.!
-;;
.;: CI
Tube ~ 1020- 1820
W
o
o
.:::
u '0 '0 on
c
Wedoes
.2 W
c:> on
Concrete
o
.:::. 0.1
cCI c
a.
Soil in place
Shell 01800 {maxi thickness 20mm
Figure 4 : Driven pier with direct sealing of the support (63 to 400 kV lines)
tOft
.
I
I
\
\.~
Al
:
c
0
I
a x b
I
\
Figure 5:
Principle of calculation of a shallow foundation block with base cross-sectional area a x b (a > b).
OVERHEAD TRANSMISSION NETWORK
33
pD (CM + D (Mcp+ M /) + Pt (Figure 5) foundagion perimeter (m) D foundation depth (m) C soil cohesion (MPa) volume weights of soils (N/m3) ~! and (Mcp+ M /) : coefficients depending on the internal friction c angle of the soil and on the ratio D/R, R being the radius of a circular slab having the same perimeter as the rectangular
Q ft - P
-/
-
slab (R = a ; b) [6J ' [7J
- Pt
weight of the foundation
.
and of soils plumb with the slab
(N).
312. Deep foundations on piles As indicated above, the utilization of such foundations is becoming increasingly important in France. This utilization coincides moreover with the use of a simple, recently developed design method [2J ' which is based on the pressuremeter test and mainly on the measurement of the limit pressure of the ground at several levels. This method is essentially an experimental one, since based on the results provided by several hundreds of tests ; it suggests, for different kinds of piles and in different soils, a correlation between the pressuremeter limit pressure and limit unit skin friction (Table I and Figure 6 according to [2J ). The design of the pile is then immediate. This kind of design, matched with safety coefficients given in paragraph 4, is used and wholly satisfactory.
32. Calculation of foundations for poles, subjected to overturning For deep foundations of single member supports, a calculation method inspired by the work carried out by E.P.R.I. [3J in this field, has recently been developed. Its basic principles are the following - use of the pressuremeter test through its main data : ~enard Modulus and Limit Pressure ; - design of reaction modulus (kh), by means of the formula proposed by ~!enard [4J ; the foundation block is supported on lateral springs with non-linear plastic behaviour. Lateral friction and under-base reactions exhibit, on the contrary, a linear plastic behaviour. This mode of calculation was compared with 2 full-scale EDF tests and with 14 tests carried out in the USA by E.P.R.I. The results are ·..;hollyconvenient [5J (Figures 7 and 8) and very close to those obtained by E.P.R.I. 's code PADLL. A computer program has therefrom resulted, ensuring calculation of deformations of the foundation submitted to given applied moment ; search for optimum installation depth, by using of displacement and rotation criteria provided by the user ; - design of the moment-displacement curve at the ground level and the analysis of the limit overturning moment.
34
TRANSMISSION LINE TOWERS FOUNDATIONS TABLE I : Pressuremetric rules Table for the selection of nomographs (Figure 6) relative to unit friction (according t~
B
AC bis
5 APa) bis F
A (A) (C) - *-bis ---(C)
15- to 40
*
* **
*** F F * * *** A A > 10 F F E F > pressure E F >> E D > AMetal C B F bis bis Aody bis10 < B AE 725 7 B * D** D Concrete Low Tubed Limit drilled Driven-in F F 20 to 40 20 Implementation A bis AA A bis pressure Pressuremeter Concrete A bis (C) B bis (B) (A) (A) (B) (A) (A) (C) (B) (A) body (C) (E) High body body b pressure Injected **A bis bis *** > 45 bis Drilled IAAB AAConcrete bis
and nature of the pile body A bis I
to silty or (10 Type wea weathered Marl thered andted mar to Fragmen
*
The values
in parenthesis
(
) correspond,
of the pier and to an implementation
for drilled
technology
piers,
implying
soil in contact with the pier body, but for the driven piers .
**
soil around
the pier •
Recommended
for soils whose
*** Only for cases where driving
5 pl > 15 10
Pa
is possible.
to a careful
minimum
execution
rehandling
to a tightening
of the of the
OVERHEAD
1.2 1.8
~
0
OJ
5
2 '"
1.41 1.6 Cl. c .9 ~ug 1.0 0.6 Cl. :J 0.8 ~ S2 '0 :§ ·c u 2.2 :g.§ 0.4 6.0 0.2 2.0 --1
on on
10 _
oV
1-
,
q sI (105 Po)
TRANSMISSION
35
NETWORK
,
I
o
Qs (1Q5po)
c B
A
A bis
20
25
30
Pf (105 PO)
F
0
_~
'
:J ,(\5pO I I.II E! 4.0_(\10~1 ~ --1
.~
---L--1
E
2.0
o o
10
20
30
Limit pressure Figure 6
40
50
(pressuremeter
test)
New pressuremetric rules a) nomographs for the determination of the unit friction q (clays, loams, sands, gravel, chalk, marls, organic s soils) b) nomographs for the determination of the unit friction q s (very compact marls, rocks)
The criteria to be adopted to use this program, is under consideration. Presently, a verification of the pressure induced at all the ground concerned levels, associated with a limitation of the displacement to 1 cm under working loads is the rule.
36
TRANSMISSION Figure
7
LINE TOWERS
FOUNDA nONS
Comparison between measured displacements
and calculated upper
line x=y lines y= 2x and y = O,5x mean line
•
C:'L':'!
S
o ...·;.iHN
Figure 8
Comparison between measured
and calculated
/
,/ /
c
£?~ j
7
T
E?R;
12
J..
£??J
13
v
E??!
}'
.:.
£?r.;
11
upper rotations.
line x = y lines y= 2x crd y=O,5x meon line
, •
o.~
'.>
Meosured
rotation~'
(degrees)
c:..:...:..!S
OVERHEAD 4. Safety factors 41. Shallow foundations 411. Compression
TRANSMISSION
for four-legged
NETWORK
37
lattice towers
The stress taken into acount in compression is that resulting from the application of a ice-loading hypothesis (2 or 4 cm thickness of rime whose density is 600 kg/m3) : it will be verified that the resulting compression stresses under the foundations are at most equal to the maximum permissible punching stress divided by three and this whatever the kind of loading, permanent or not. 412. Tension The tensile stress taken into acount is the highest one resulting either from the application of the "administrative hypothesis" (wind of 110 km/h) multiplied by 1.5 or of an assymetrical ice-loading hypothesis (2cm/Ocm or 4cm/2cm). Two cases are considered : 1°) If towers angles are less than 30 degrees, the relation between the ultimate load calculated Qft and the higher of the two stresses indicated above should be greater than 1. 2°) In case of "dead end" towers or angles greater than 30° the relation between the calculated ultimate load Qft and the higher of the two stresses indicated above should be greater than 1.2. 42. Deep foundations for four-legged lattice towers We will merely indicate the rules adopted for uplift. For compression the permissible peak resistance to take eventually into consideration is affected by the coefficient 3 with respect to rupture as in the case of shallow foundations. For uplift the stresses taken into account are chosen under the same conditions as in the case of shallow foundations, two possibilities may arise : 1°) If the tower angles are less than the ultimate stress calculated Qft greater than 1.4. 2°) If the tower angle is greater than the calculated ultimate stress Qft greater than 1.7.
30 degrees, the relation between and the stress applied should be 30 degrees, the relation between and the applied stress should be
As indicated in paragraph 3, notion of safety deep foundations subjected to overturning.
factor is not used
for
38
TRANSMISSION LINE TOWERS FOUNDATIONS Bibliography
BIARREZ - BARRAUD Calcul des fondations superficielles a dalle foundations with base plates) Paper 22106 CIGRE Session 1968 (in English)
(Design of shallow
BUSTAMENTE Michel - GIANESELLI Luigi Prevision de la capacite portante des pieux isoles sous charge verticale (Prediction of bearing capacity of separate piles subjected to vertical load) Bulletin de liaison of LCPC nO 113 Hai-Suin 1981 (in French) E.P.R.I. Design of laterally loaded drilled pier foundation Paper n° EL 2197 January 1982 [4 J M. GAMBIN Calculation of foundations subjected to horizontal pressuremeter data SOLS/SOILS n° 30/31 1979 (in English)
forces using
[5J J.L. LAPEYRE - M. GAGNEUX - J. VIEILLE Calcul des fondations de supports de lignes aeriennessoumises renversement : deux besoins differents et deux approches differentes (Calculation of overhead line support foundations subjected to overturning : two different requirements and two different approaches) SEE Symposium on "Foundations" 27 November 1986 (in French)
au
[6J MARTIN Daniel Calcul des pieux et des fondations a dalle (The design of piers and pad and chimney foundations) Annales of ITBTP n° 307/308 July 1973 (in French) [7) MARTIN Daniel
- PORCHERON Yves Etude de la rupture des fondations de pylones sollicites a l'arrachement (Study of the rupture of tower foundations subjected to uplift loads) Bulletin of Direction des Etudes et Recherches (E.D.F.) June 1968 (in French)
[8J TRAN-VO-NHIEM Force portante limite des fondations superficielles et resistance maximale a l'arrachement des ancrages (Limit bearing capacity of shallow foundations and maximum uplift resistance of anchors). Thesis for Doctor Engineer degree, Grenoble University, 12 February 1971 (in French).
STEEL PILE FOUNDATIONS FOR TRANSMISSION LINE TOWERS, AS USED IN WESTERN EUROPE. Alexander J. Verstraeten (1)
INTRODUCTION. The design and construction of foundations for power transmission line towers present some special problems. This paper describes a system of design and construction for these foundations that was developed in Europe and has attained general use there because of its reliability and cost effectiveness. Transmission line foundations distinguish themselves by having to deal not only with compressive and lateral loads, but with uplift loads and, because of the wind (in some regions earthquakes), with dynamic loading. The construction of transmission line foundations distinguishes itself mainly in the great number of inaccesible locations involved, resulting in extensive geotechnical investigation and logistical problems in moving men, materials and equipment. This paper describes: - A foundation system for lattice and single pole tower transmission lines using steel pipe, prefabricated piles. The system allows much of the work to be shifted form the fie ld to the manufacturing plant, speeds up the work in the field, and creates highly reliable foundations. - The Cone Penetration Test (CPT) based design method for transmission line foundations developed by the Delft Soil Mechanics Laboratory of the Ne therlands. We generally consider the CPT, where applicable, to be the most reliable and cost effective method of geotechnical investigation available. The CPT based design method reduces costs by allowing the optimum choice of pile type and length of pile, and by minimizing "surprises" in the field. - The methods and equipment transmission line foundations.
used
in
installing
the
HISTORY Before describing the technology of the pipe-based .transmission line foundations we will discuss the historical factors that led to its development. The first foundations for Dutch (lattice) transmission towers consisted of 4 groups of timber piles, driven through very soft
1
President, Fundex Companies, The Netherlands
P.O. Box 55, 4500 AB Oostburg,
39
TRANSMISSION
40
LINE TOWERS
FOUNDATIONS
and compressible toplayers into a dense sand stratum, and capped by reinforced concrete caps, which were in turn connected by large, reinforced concrete beams. The timber piles carried compressive loads only; the weight of the foundation caps and beams carried the uplift forces. The next development was that the wooden piles were replaced by reinforced concrete piles. Because of the larger bearing capacity of the concrete piles it was possible to employ fewer piles. The concrete piles were also able to withstand moderate uplift forces, so that it was possible to reduce the weight of the pilecaps and connecting beams. At a limited number of locations with sandy soils of larger bearing capaci ties, sha 11 ow foundations were used. These consisted of concrete foundation blocks; in order to save on concrete the blocks were prestressed onto the soil by means of almost vertical placed, grouted tie-backs. The application of this type of foundation was limited because in most cases it proved to be less competitive than pile foundations. Further simplification was achieved with the introduction of the pipe-based pile system, which has now become the most common system in use. The pipe-piles can carry such large uplift forces that the application of a single pile per tower leg is possible, pilecaps have become unnecessary, and the connection between piles and legs has become very simple. The system has resulted in foundations that are highly reliable and cost-effective, and that can be installed in a minimum of time. SOIL
CONDITIONS.
The develpment of the pipe based foundation by Dutch soil conditions.
system was influenced
Dutch soil generally consists of a layer considerable depth of young deposits without any sound rock on which to base a foundation. The soft top layer can run to a depth of up to 65 feet and, because of consolidation, subsides at a rate of up to 1 foot per 100 years. All pile foundations are driven 6 to 10 feet through this soft layer into the bearing sand stratum underneath, and are therefore end bearing. The groundwater level is usually feet below the ground surface. Furthermore, more densely
high
and not more
than 1 to 3
the accessibility of the low lying polders in the populated Western part of the country is poor.
In order to resist the large uplift forces that are exerted on transmission towers, a deeper than usual penetration into the sand stratum is required, resul ting in hard driving conditions. Steel pipe piles are very suitable under such circumstances.
STEEL PILE FOUNDATIONS The steel pipe piles generate large side friction as well as large end-bearing. They are also very suitable to transfer lateral loads onto the subsoil. Their relatively light weight and large strength make them attractive for transport under difficult circumstances. In areas with poor accessibility pipe-piles have been transported by helicopter and piling machines have been moved from mast to mast location, using specially made hardwood 20'x3' movable mattrasses that spread the machines weight over a sufficiently large area. Because precast concrete piles take up 80% of the very competitive piling market and are manufactured industrially, their prices are low and a supp ly from stock is norma 1. Stee 1 piles are more expensive. However, because of the aforementioned reasons the application of pipe-piles is more economical than that of precast concrete piles. THE
STEEL-PIPE
PILES.
For lattice towers the most commonly used pipe-piles are closed ended; the closed end compacts the soil and improves performance. However, where higher frictional forces are required the outside of the closed ended pile is provided with a groutmantle. The groutmantle increases the bonding with the soil and therefore pile performance. wnere hard substrata cause undesirable driving rriction an open-ended pile with outside and inside grouting can be used. During driving the grout reduces rriction and stops plugging; after driving the outside grout improves bonding with the soil and the inside grout acts as a plug. For single- and double pole towers only open-ended, wide-diameter pipe-piles are used. The closed-end
pil~
The closed-end pile (without groutmantle) is shown in figure 1. The closed-end pile is generally used for tensile loads of 50 to 60 metric tons, and compressive loads of 90 to 120 metric tons. Diameters range from 355 to 457 mm (14" to 18"). The soil displacement caused by the closed end improves the pile's performance; the pointed shape of the closed end further improves performance.(see further below) The tower and the pipe pile are connected by a stub that is ancred by 6000 psi concrete in the top of the pipe. To achieve sufficient bonding capacity to withstand shearforces ribs are welded on the stud and inside the pipe. The stub and pipe ribs are staggered, with the lowest rib on the stub placed well below the lowest rib in the pipe. See rig. 1. (The carrying capacity of the bond between the (almost) vertical surfaces of the concrete core and the inside of the pile, as well as the (almost) vertical surface of the stub and the concrete are ignored in practice).
41
42
TRANSMISSION LINE TOWERS FOUNDATIONS
;.5' - 50'
.
." .
....
,
.
' ,
.~-'~--':'-'.'_' . Grav~: or '--S,3n·j ._-------
/
(oncr •.le
Fig.1
ring
/ Coner('!('
FIg- 2.
Groui
2.5"-)"
\. '. Nak('d
surtae~
to oel 05 C sat ••ty earth
43
STEEL PILE FOUNDATIONS Each rib of the stub is considered
to load the concrete over
its
(almost) horizontal surface, while the rings inside the pipe take over the same load from the concrete. The concrete core itself is loaded in shear. The shear surface is taken as the distance between the lowest ring inside mul tip 1 ied by the circumference
the pile and the pile top, of the rings ins ide the pipe.
The allowable shear stress is normally 7,6 kg/cm sq. = 106 lbs/sq.inch. The allowable compression on the ringsurface is 110 kg/cm sq = 1500 lbs/sq.inch. To prevent corrosion, the outside of the top-end of the pipe is shotblasted and coated with epoxy resins down to 3 feet below the water table. Corrosion at deeper levels can be ignored. After installation the pile is filled, up to 5' below the with clean sand, gravel or lean concrete. Next, the stubs mounting the tower leg are fixed in position by tack-welding to s trip s that are we 1de d tot he pip e. Th is a 1sot ak e s car grounding the tower. Finally, the top 5' of the pile concreted. The "closed-end,
grout-mantle
top, for them e 0f is
Ei.l~
The "closed-end, grout-mantle pile" is shown in figure 2. For frictional forces in excess of 50 to 80 metric tons per pile, increased capacity is obtained by providing the closed-end pile '",itha 2 1/2 to 3 1/8 inch grout-mantle. The grout-mantle provides improved bonding to the soil. The pipe diameters used for the "closed-end grout-mantle pile" and the corresponding pile bearing capacities are as follows: 60 - 120 mt: Maximum uplift 120 - 160 mt: Maximum uplift 160 - 250 mt: Maximum uplift
dia. 609 mm dia. 762 mm dia. 914 mm
24" 30" 36"
To make the grout-mantle a collar is welded just above the point of the pipe and grout-hoses are placed from the top of the pipe to the collar. During driving grout is pumped through the hoses to the anular space that the collar creates. The grout used for the mantle is a mix of 550 kgs cement, 1200 kgs sand, water and an additive to keep the mix sufficiently fluid and pumpable. Very often the additive Tricosal is added to reduce shrinkage during hardening. Practice has shown that at the start of groutpumping the required pressure is low. At penetrations over 50 feet pressure will have to be increased significantly, mainly because the mortar in the top part of the mantle has dewatered so much that it has lost its fluidity and prevents the upward escape of freshly pumped material from lower levels. When the pile has reached its required penetration, pressure is increased to appro 10 ato. (15 psi). Application of this additional pressure further
44
TRANSMISSION
LINE TOWERS FOUNDATIONS
improves bonding with the soil and pile performance by forcing excess water into the surrounding soil while the grout densifies and stiffens. The amount of grout theoretical volume. of course less than grout level should possible. However, prevented.
required runs from 1.1 to 1.5 times of the In the deeper layers the over-consumption is at the top. To avoid excess use of grout the be maintained as close to ground level as some excess upward flow can usually not be
As the grout mantle bonds very well to the steel of the pile, it protects the steel surface against corrosion. However, the top of the mantle is removed to 4" below the ground surface and the steel surface is coated with bitumen or epoxy. The short ungrouted pipe point extending below the collar acts as a guide during the early stages of driving and, after completion of the tower, guarantees grounding. Open-ended
pile with inside and outside grout-mantle.
Wnere the substratum is particularly hard and difficult to drive into it may be advisable to use an open-ended pipe and grout both the inside and outside. The open-ended pipe will reduce total soil displacement (as compared to a closed-end pipe) and the inside grouting will reduce friction and prevent plugging during driving. For inside and outside grouting a minimum diameter pipe of 609 mm (24") is required. Collars are welded inside and outside of the pipe and a number of holes are made in the pipe-wall to allow the grout to move freely from the outside to the inside anular spaces. After hardening of the mantles, the bonding of the inner groutmantle guarantees a "plugged" behaviour under service conditions. After the inside plug is augured to a depth the inside is cleaned, the pile is completed as the other two types of pile. Tube foundations
for single- and double-pole
For single- or double-pole diameter are used.
towers hollow
of 6 to 8 feet and in the same manner
towers.
pipes
of up to 8 feet
After the pipe-pile has been driven the soil core is removed to a depth of 8 to la' and the pile's inside is cleaned. On top of the remaining soil plug a base slab of lean concrete is cast. Next, a steel plate with a conical pin in its center is ancred wi th concrete in the middle of the s lab. See fig. 8.
45
STEEL PILE FOUNDATIONS
,.,.---1--- ~~'---~ --":f~----
----Conc-.rete Steel wedses
Clam~_2.lece
/
I
% I I I
~-=Conical
,
Cencrete
/
.) ,..:"
/. 0' ~
pin
, I
/ Anchor ~, '-(steel)
~.I'
,
Clamp piece
-----~~_.
'-
Fig 8a.
Concrete
9cse ~la!LQL
,'.'" ;..;/ ..
, 'd:>
leon concrete
Fig 8.
Steel
:, '''','"", ! ~~,~1 .;~ ... t.~1'? ~ i ~~~1)'"" ' .""",""-
..... -- :,\,' ..., ;.. ~ --'" ~ •. '¥ ~t ,~', jI s! _.' ,J I" ' Jto "-"',I!)f 7.-I "J' , ' .. , ~ { '" ( ;:..'" :5.~·'f "-An~or /,r/' ~ Fig 111")8b.'v ..• I)' '-'," ~. "J' '.
-
J
'·.·
'../
••
1
'\,
',,,
bol t s
"-
flange
46
TRANSMISSION LINE TOWERS FOUNDA nONS
In the center of the foot of the mast a hole has been made, that matches the conical pin. The mast is placed in the pipe-pile and centred by placing the hole over the pin. The mast is positioned vertically by means of steel wedges placed between pipe and mast. Next the space between pipe-pile and mastfoot is concreted. See Figure 8a. Another possibility is to equip the polefoot with a thick hoizontal steel flange with a number of anchor holes. Prior to placing the mast the pilecore is concreted while a matching number of long anchor bo 1 ts is placed and he 1 d in the ir exac t position. See Fig. 8b.
_DE_S_I_G_N_I_N_G _F_OU_N_D_A_T_I_O_~_S _FO_R_P_R_EV_A_I_L_I_N_G _SO_I_L _C_ON_'D_I_T_I_O_N_S _AN_TD _NA_T_U_Ri_A._L FORCES In order to design powerline foundations that will meet requirements at minimum cost it is essential that thorough geotechnical investigations be carried out. Experience in the Netherlands suggests that it pays to carry out investigations at all tower locations. The distances between the towers, varying between 1000 to 1600', are so large that substantial variations in the soil profile may occur between locations. These variations must be known beforehand to allow the pipe-piles to be premanufactured at their optimum length, and to prevent unnecessary interruptions in the field because soil conditions turn out to be different than expected. For Dutch conditions the static Cone Penetration Test (CPT) is the most effective soil investigation method (the Standard Penetration Test (SPI) is not used at all in the Netherlands). The Cone Penetration
Test (CPT)
For those not familiar with the CPT, it can best be described as a miniaturized and instrumented model pile (the cone) that is pushed into the soil while the end resistance and the side friction of the cone are measured and recorded in relation to depth. Ihe depth to which depth that the piles will
the cone is pushed reach.
CPT readings are made for every gaurantees that even very thin soil
is greater
than the
inch of penetration. This layers do not go unnoticed.
During its penetration the cone displaces the soil. Its behaviour is comparable to that of a displacement pile and CPT readings are therefore predictive of the bearing capacity of displacement piles. The end resistance, as measured with the CPT, must be scaled up in order to arrive at the correct end-resistance for an actual pile. The skin friction has been shown to be independent of the pile size and can thus be applied directly.
STEEL PILE FOUNDATIONS CPT's have also proven to be re 1 iab le indicators of soi 1 type. CPT soil type analysis is done on the basis of the so-called measured friction ratio, that is: cone-resistance divided by local friction, times 100%. See fig. 4. Usually a soil-type analysis based on CPT readings is far more reliable than the description of the soil profile given by a drillforeman. A further advantage of the CPT is that results are independent of the skill and experience of the operator; if 2 operators perform a CPT at the same location the same results are obtained. The analysis of CPT data is increasingly being facilitated by the computer. Usually CPT readings are recorded on tapes or discs and later processed by a computer which will plot the cone resistance and the skin friction in relation to depth. See fig. 5. Software is available that will plot the pile's allowable bearing capacity as a function of depth (provided data on the applicable safety factor, pile size and pile type are entered). Increasingly computers are operational in the field and process CPT readings in real- time. The main disadvantage of the CPT is that in some soil conditions the cone will not penetrate to the required depth. The pushing capacity of the heaviest CPT equipment is 20 tonnes for a standard cone and rod system of 36 mm diameter. This force is sufficient to push the cone with rods through shales, marls and other soft rocks. Soil containing sound rock and larger sized stones make it necessary to combine CPT's with drilling techniques. Another disadvantage is that because CPT's are only now becoming generally used in the US, the available data from the past are mostly SPT data. This may require conversion of old SPT data to allow comparison with new CPT data. Nevertheless we believe that, where applicable, the CPT method is the mos t cos t- effec ti ve geo technica 1 inves tigation technique available. It is relevant to mention here that Larry Nottingham of the University of Florida did extensive research into the capacity of different methods for predicting the bearing capacity of a number of piles used and tested in the USA (Doctoral Dissertation 1977). The work was done under supervision of Professor John Schmertmann. One of the systems investigated by Nottingham was a CPT based system developed by the Delft Soil Mechanics Laboratory of the Netherlands. (This system is described below). Nottingham came to the conclusion that he could not improve on the CPT/Delft method; it came out as the clear winner. In the Netherlands the confidence in the Delft method is so complete that less than 5 pile load tests per year are done to check on actual bearing capacity, remarkeable for a country where pile foundations are used on a larger scale than in any other country of the world.
47
TRANSMISSION LINE TOWERS FOUNDATIONS
48 MN/m'
8c
40
ro
'" ;;.
1.3
~ OJ
C
ou 30
20
10
o
o
0,3
0,2
0,1
-
Fig
L..
Relation soil for
local
between the friction ratio and the mechanical adhesion Jacket
ConE'
":'5IstaneE'
100 E
II
c
,
in kg/em2
IfictlOn
the type of cone
_ 30e
200
I
I
.!:
0.
o~
--
20 AeeumulatE'd _"-
Loeal f rI e t Ion In kg/em2
1000
E
:::. 1
o'" I ,
15
10 Fig.5.
frictIon .•.,
2000
ko/cm -..• c"'-"r~urn fE"E'nc E'
3000
STEEL PILE FOUNDATIONS Because of its superior predictive capabilities the CPT techno logy allows for the optimum des ign of foundations, resulting in savings on materials and improved productivity in the fie 1d. Pile Design The Delft Soil Mechanics Laboratory of the Netherlands has done extensive research into the correlation of CPT data and the actual bearing capacity of different types of displacement piles. This has resulted in reliable design procedures for displacement pile foundations, including foundations for transmission towers. Most piles only undergo static compression loads and pile penetration is determined on the basis of tensile capacity generated under static conditions. Since friction under tension is equal to friction under compression, the maximum compressive capacity of closed-end piles is calculated by adding the end resistance to the total friction resistance. However, the factors determining foundation design for lattice transmission towers are; - the dynamic pattern of uplift and compression forces, which in turn depend on, - the position of the tower in the line, - the natural forces exerted on towers and cables, - and the weight of towers and cables, - the strength, flexibility and shape of the piles, the volume of soil displacement and the form of the pilefoot, and the effect of such techniques as pressure-grouting of the pi le pipe. Transmission line towers can be positioned in three ways on the line. There are the dead-end towers, which are longitudinally loaded from one side by the suspension cables, (the overturning moment acting on such towers is large), the tangent towers, placed where the line makes an angle, which are vertically as well as transversally loaded, and the towers on a straight line between two other towers, which under static conditions carry a vertical load only. Normally the design load per footing for suspension towers varies between 20 and 60 mt in uplift and compression. For both other types of towers, the pileloads may vary between 50 and 250 mt in uplift as well as in compression. As a result of the position of a tower in the transmission line and the dynamic effect of natural forces different patterns of dynamic loading of the foundation result. In general, these can be divided into 4 types (see fig. 6.): I. The load alternates between compression and uplift. II. The load alternates between small and large uplift. III. The load alternates between the maximum uplift and zero. IV. The uplift is constant.
49
50
TRANSMISSION LINE TOWERS FOUNDATIONS
I', + :'.
•
+' ~, ; I
"
! i Ii! ·1-
!,
-
I'
P.~
:.
...!...l..L
iime>~
I
".
I I , I I '"
iime>
II
..
-..
i i me> ---...
Case Decrease
of friclion
I : 0,35 II: 0,50 III: 050 IV: 100
Fig.6.
Accumulated
fnclicr
Ii I~
~
I
5
l.
b
kg/cm
Clrumference
2000
3OJO
I
-~
f'\r---1..
I r
~ !
!
~ 10
15
WI Pi Ie point level
i
Y-
I I
I,
20
::
I
Fig. 7.
_
STEEL PILE FOUNDATIONS
51
Dynamic loading results in continuous pile movement that causes deterioration of frictional bearing capacity and relatively large foot settlements. Experiments have shown that the effect of this deterioration is concentrated in the middle section of the pile and varies with the type of loading pattern. Figure 6. graphically illustrates the different loading patterns; next to Decrease of Friction are given the factors indicating the effective friction in the middle of a pile that undergoes the corresponding type of dynamic loading pattern. The greatest deterioration of friction results where the load alternates between tension and compression, such as illustrated for type I. As mentioned before, the shape of the pile, the volume of soil displacement, the form of the pilefoot, and such techniques as grouting, also influence pile bearing capacity. This influence has been experimentally quantified by the Delft Soil Mechanics Laboratory in a "factor p", for which some values are as follows: for piles with a flat underside or open pipe piles and H-beams: factor p = 0,30. for piles with a sharply pointed foot: factor p = 0,55. for open pipe piles with an injected outer mortar mantle: factor p = 0,80. for closed ended pipe piles (flat shape) and an injected outer mortar mantle: factor p = 0,95. Numerous field tests have shown that it is very advantageous to equip pipe-piles with a mortar mantle, as their total skin friction is almost three times as large as that of ungrouted pipe piles. The Delft Soil Mechanics Laboraratory has also found that prefabricated piles (steel or concrete) with a pointed foot can generate almost twice the side friction of piles with a flat foot (but tend). In view of the above, the De 1 ft Laboratory has deve loped following method for calculating pile length.
the
Based on experience a certain pile length is assumed. This length is divided into 3 parts for each of which friction will be calculated seperately. These parts are;
minus
A. a top part consisting of the top 1/4 of the pile-length the top 1 meter, B. amiddle part; being the next 1/2 of thepile-Iength, and, C. a lower part; being the rest 1/4 of the pile.
The friction measured each part of the pile The to ta 1 friction follows: The sum of:
by the CPT for the corresponding is totaled (see figure 7). capacity
of the pile
depth
is calculated
of
as
Total (CPT)friction part A, Total(CPT)friction part B, mul tip 1 ied by the appropriate "decrease of friction factor",
TRANSMISSION
52
LINE TOWERS
FOUNDATIONS
Total (CPT)friction part C, Mul tiplied by: The circumference of the pile, Multiplied by: The factor p.
~Q!:
£~lCUl~!lQ~ Q! !Q~ ~Ellft £~E~£l!y Q! !he Ell~ th~ to!~l £~E~£l!Y Q! !Q~ Ell~ l~ ~Qi~~!~Q E.YQlYlQl~g E.Ya
friction
safety factor ~
which is usually
taken to be between ~ and ~
Once pile bearing capacity has been established for the asssumed pile length, optimum pile length is determined through an iterative process that matches pile bearing capacities for different length piles with bearing requirements. It will be clear that design calculations are usually computerized. Designing Single Poles For single poles the diameter of the pipe-piles is in the range of 30" to 100". The wall thickness of the pipes varies between 0,8 to 1,0% of the outer diameter. To resist extreme bending that can take place under special conditions thicker pile walls can be app 1 ied . Pile design is usually based on the assumption that the soil renders a lateral purely elastic support. This approach requires data on the spring constants of the various soil layers, which are derived from the site investigations. Computer programs are available for the determina~ion of pile strength and pile deflections for any multi-layered soil profile. INSTALLA..TION. In order to gain the full benefits of working with prefabricated pipe-piles it is essential that the piling rig can be mobilized, transported and demobilized in a very short time. For this purpose Fundex Piling Equipment B.V.of the Netherlands has designed and built the Fundex rig with fixed guides that is very easily mounted and dismounted in the field. Transport from tower to tower location is usually done per low-loader because this is faster than having the machine move by itself; rarely is it possible to follow the shortest route between towers. Lattice tower pipe-piles are driven under an inclination that matches that of the tower legs. The Fundex rig is constructed to do this. For single pole pipe piles pile installation is easier because the position of the pile is always vertical. Both impact hammer and vibratory hammers are used; in cohesive soils the impact hammer is more effective, but in saturated granular soils vibratory driving can be very effective. For purposes of driving the Fundex rig is usua lly equipped wi th a diese 1 hammer of the Delmag D-30 type, which supplies 80,000 LBF on impact. The Fundex rig has made foundations in a day.
it possible
to install
several
tower
STEEL PILE FOUNDATIONS
53
TESTLOADING. Testloading under tension is relatively simple just as is loading horizontally. It is more difficult and expensive to test under compressive load, because this requires either a large dead we igh tor a suffic ient ly 1 arge numbe r 0 f grouted tie -backs to supply the large reaction force required. However, we are very aware of the fact that test loading is always necessary to gain sufficient insight and confidence in a new foundation system. If such testing is done under the guidance of an expert, the program can be limited to the essentials and the cost and time loss minimized. In the Netherlands the contractor usually sets up the test and an expert engineer or consultant, such as the Delft Laboratory, carries out the test. we intend to do tests of the pipe-pile system for transmission 1 ine foundations in the USA and make arrangements for American experts and consultants to carry out these tests. Over the years we have developed practical and effective test procedures to establish or extrapolate failure load for the powerline pipe-piles. The procedure is to do anumber of compression/decompression loading cycles at increasing loads while registering the uplift after every cycle. The cycle loads are increased in increments of 10 to 12 1/2 % of the projected failure load. The first cycle at a particular load is maintained for some time to establish time/settlement behaviour. The next 4 cycles are short, after which another series of cycles starts at a higher load (see fig.9). It has been experimentally established that when one of the compression/decompression cycles at a particular load results in a rise of 0,2 mm, the respective load is very near 50% of the failure load, which is usually also considered the maximum allowable design load. This knowledge is important in cases where it is impossible to load the pile to failure. where it is possible to load the pile to failure, the failure load is established when there is cumulative permanent rise of the piletop of 20 mm or more after a cycle. Lateral deflections under the maximum allowable should remain within the elastic range.
horizontal
load
CONCLUSION. The design and construction of foundations for power transmission line towers presents some special problems. These foundations have to deal not only with compressive and lateral loads, but with uplift loads and, because of the wind (in some regions earthquakes), with dynamic loading. The large number of inaccesible locations involved result in extensive geotechnical investigation and logistical problems in moving men, materials and equipment.
54
TRANSMISSION LINE TOWERS FOUNDATIONS
fT
1)
II
----·--·10·
up
..·
-
D
i
I
/1 10
,~!
I
b '/
,
• I
) I
I!
I
I
II·
i
I I
)1
.s .c
1'- I ,,! III!
c.
J/
o•.
I
I
. ~
,
~.
.
I il
I
I~
I I
1-.;::'
Tub•• 91~mm,(36")
===fll i I
15
o
I
I
I
I '_ ~:
II
II
I
I
I
I
, I
i iI
I
,
I
I I
I!
-n .xI
iI
I
I
i
I I I
I
I
Ir--~!
l000psl
Dutch
-.
~
n Ii
,
I
I
I
,
I! I
I
I! 1000 psi
J
I
I
I
I
I
I
I
I
3000 psi
Cone Penetration Test for test pile ~ 36"
j.
z11500
"
13lOKN 1050KN
" 1000 .. o
-: 50 Oi ..
•..
o
1 T Im~ in hours
1 --.
3
5
TIME - LOADDIAGRAM
1//8
. o
5
Tlmf' in hour!t ~
ig9:Upllft
tesllood on a groutinJe:tedpile
¢36'
STEEL PILE FOUNDATIONS
Ii,IPI! CI : I
I , ~I !I Ii \ II -i(I<: I ,,I ':>i, I Ii i !! ili
I
I
I --1:~
0
, J
!
o
-
COl"M!'r"sislan" in , kg/cm1 II I', I I '0 , 3({ IIiII!i!• . II!,II!IIII' ' I,:!1III,!I.,,i!I!iI, II!!,iI!IiII Ii: i2aJ IIi!II•:I I IiII!IIjI ! Ii I !
I I I, I I
_
55
Local friction o
1
In kg/cmZ 3
Z
__
,
1
I
i I
-5 .c Local
::>.
a..
!rlellon
-10
-15
o
1000 zero
!XC
--- - ....
frlCllon
Cummu1alpCl
'000 ltg/em
radII r~56' ... lForce~
IS
II,
!3 M",asuring-rod
,radII
I
11
-J._=- ..
tl210
---r-r-i i~
Eg
Measurinc-rod
Ea
c
II
.- 7
~c
E
6
J50KN ~ __
--r--
5
~ I,
~...J a. ~ 2
°1
50KN
o
5
L 10
15
20
25
))
Time in minutes rig9a:
Horizontal
test load on' pile
35
1,5
~ 0;6"xS8"wallthlc~ness
55
56
TRANSMISSION LINE TOWERS FOUNDATIONS
Because of its reliability and costeffectiveness the pipe-based foundation has gained a large share of the European market for transmission line foundations. For fast installation of pipe-piles with a safe holding capacity of up to 80 mt, plain steel pipes with a pointed, closed end are competitive and fast to install. The bearing capacity of the closed-end pipe-piles can be much increased by injecting an outside grout mantle during driving, or for large diameter piles, by doing this both at the in- and outside of an open-ended pipe. Grouting facilitates pile installation and gives a reliable protection against corrosion. The higher bearing capacity of grouted pipe-piles make it possible to apply one pile per tower leg for any type of tower structure. Single pole transmission· towers can be founded on large-diameter single pipe-piles, which are simple to install and have proved to be competitive. The pipe-pile advantages of:
foundations
for
transmission
towers
have
the
- Avoiding unnecessary earth work and field damages. - Reduction of the in situ application of concrete for pilecaps or drilled shaft piles. - Where the groundwater level lies at a short distance below the groundsurface there is no need for dewatering or danger that the quality of concrete structures suffer because of groundwater. - Simpler and consequently faster construction. - Straight forward load transfer from tower leg to foundation pile. The CPT based design system developed by the Delft Soil Mechanics Laboratory has allowed the optimization of transmission line foundation design, allowing for a smaller design safety factor, and minimizing costs and "surprises" during installation. Penetration depth of these piles is usually determined by the maximum uplift force and not by the maximum compression. The Fundex rig which has been designed for installing pipe foundation systems for transmission lines; it provides very short mobilisation and demobilization times, easy transportability, and the capacity to speedily install pipe-piles with the required accuracy and at the required angle. Only positive experience has been gained with thousands of pipepile foundations for transmission towers in Holland, Belgium, Western-Germany and France.
Uplift Braja
Capacity ~!. Das,l
of Model M. ASCE,
Group
Anchors
and Yang
ill Sand
Jin-Kaun2
Abstract Small-scale laboratory experimental results for the ultimate uplift capacity of shallow horizontal circular single and group anchors embedded in sand have been presented. The experimental ultimate uplift capacity of single anchors has been compared with theories provided by Meyerhof and Adams (7), Vesic (8), and Clemence and Veesacrt (~). For anchor groups, the uplift efficiency varies with the number of anchors, center-to-center anchor spacing, embedment ratio, and soil friction angle. The experimental uplift efficiency of group anchors has been compared with the theory of ~leyerhof and Adams (7). Introduction Horizontal anchors are often used in construction of foundations such as transmission towers to resist vertical uplifting forces. During the past 15-20 years, the results of several investigations (both theoretical and experimental) related to the ultimate uplift capacity of single anchors embedded in sand have been published. Important contributions in this aspect can be found in the works of Adams and Hayes (1), Baker and Kondner (2), Balla (3), Das and Jones (5), Esquivel-Diaz (6), ~{eyerhof and Adams (7), Vesic (8), and Clemence and Veeseart (4). Vesic (8) has provided a review of most of the important works on this topic. In many cases however horizontal anchors are used in groups. Until this time, only a limited number of studies relating to the uplift capacity and efficiency of horizontal group anchors have been published. The purpose of this paper is to report some laboratory model test results of shallow group horizontal anchors in sand. Immediate practical application of the results obtained from this study may be somewhat limited, primarily because of the fact that many of the present transmission lines have guy tensions far greater than what a shallow group anchor
Iprofessor, Department of Civil Engineering, The University of Texas at El Paso, EI Paso, Texas, 79968 2Graduate Student, Department of Civil Engineering, The University of Texas at EI Paso, EI Paso, Texas, 79968
57
TRANSMISSION
58
LINE TOWERS
FOUNDA TrONS
would support. However, the results SJIOW the general trend for further research in the area of uplift capacity of shallow and deep anchor groups. Uplift
Capacity
of Single
Horizontal
Anchors
A review of most of the theoretical studies for evaluation of the ultimate uplift capacity of single horizontal anchors embedded in sand has been given in an excellent paper by Vesic (8). It is not the intention of this paper to review all pertinent theories; however, the theories for circular anchors provided by Vesic (8), Meyerhof and Adams (7), and Clemence and Veesaert (4) will be briefly discussed below since these are the most widely referred to in literature. The general parameters of a circular anchor embedded in sand are shown in Fig. la. The diameter of the anchor is B, and it is located at a depth D below the ground surface. If
F
q F =F*
.
;."
q
.......
q
I
Sand y
D
cp
1
I
~
... I
Shallowl anchor
Embedment ~
B
=
Deep anchor
ratio,
D/B
~
diameter
(a)
(b)
Figure 1. (a) Geometric Parameters of an Anchor in Sand; (b) ~ature of Variation of the Breakout Factor With Embedment Ratio the depth of embedment is relatively small and the anchor is subjected to a gross ultimate uplifting load Qu' the failure surface extends to the ground surface as shown in Fig. la. This is referred to as a shallow anchor. However if D is relatively large compared to the diameter B, local shear failure in soil around the anchor takes place and the failure surface does not extend to the ground surface. This is referred to as a deep anchor. The critical embedment ratio at which the transition from shallow to deep anchor condi-
MODEL GROUP ANCHORS
59
IN SAND
tion takes place depends upon the relative compaction of the soil. For loose sands (¢~300), (D/B)cr~4; and for dense sands (¢~45°), (D/B)cr~8 to 9 (7). Perhaps a better parameter for correlation of (D/B)cr would be the relative density, Dr. Figure 2 shows the nature of variation of the critical embedment ratio with relative density as obtained
----
:.:Q Q r-..'--.J U
H
68 0 4
60
40
20
Relative
density,
Figure 2. Experimental Embedment Ratio With
80 Dr
100
(%)
Variation of Critical Relative Density (5)
from the limited model tests reported by Das and Jones on square anchors. Based on their results (D / B )c r ~ 4
+
O. 0:5 3 2Dr
The net ultimate defined as
(f0 r 25 ~D r~ 7 5 %
uplift
capacity
(5)
(1)
% )
Qo of an anchor
can be
(2) where Qu=gross of the anchor
ultimate
uplift
capacity,
and Wa=self-weight
The net ultimate uplift capacity of an anchor embedded in sand can be conveniently expressed in a nondimensional form as (3) where Fg=breakout factor, unit weight of the soil
A=area
of the anchor
plate,
and y=
The general nature of variation of Fq with embedment ratio (D/B) is shown in Fig. lb. The breakout factor increases with O/B up to a maximum value Fa=Fa at D/B=(D/B)cr. For D/B~(D/B)cr' the magnitude of the breakout factor remains constant.
60
TRA.NSMISSION
Vesic's
Theory
LINE TOWERS
FOUNDATIONS
(8)
Using the principles of expansion of cavities, Vesic (8) has presented the variation of the breakout factor (Fq) with embedment ratio (D/B) and the soil friction angle (¢) for shallow circular ancho~s embedded in sand. TJlese values are shown in Fig. 3.
12
+J (J) +J
~
0-
~0;J U
•...•... roro
•.....• :::t::
846 10
2
a .5
Embedment
Figure Meyerhof
3.
Variation
and Adams'
3.5
2.5
1.5
ratio,
of Vesic's
Theory
4.5
5.5
D/B
Fq With
¢ and D/B (8)
(7)
According to this theory, the ultimate uplift a shallow circular anchor can be given as
capacity
of
(4) ,·;hereS=shape
factor=l
+
m' (D/B)
Ku=nominal uplift earth pressure coefficient, W=weight soil immediately above the anchor, and m'=shape factor efficient=f(¢) For circular
anchors
(5) of co-
,
MODEL
GROUP ANCHORS
61
IN SAND
(6)
are Eq.
The variations shown in Fig. (4) yields
of Ku and m' (which are functions 4a. Substitution of Eqs. (5) and
of (6) into 4»
0.6
1.0 (a)
Nominal uplift coefficient, Ku
/ / / /
Ku 0.8
// 0.3 m'
Shape factor // coefficient, m'/,/
/
/
, I <40 I 3Deep IIII20 45 10 40 0/ • /" /// 0.6(deg) I condition
/¢
anchor
----
/d ...•.•...
..-"""
qJ=4::J-/
,,/
5
3 o
2
4
Embedment
6
ratio,
8
10
D/B
Figure 4. (a) Variation of Ku and m' With of Fq With D/B and ¢ For Shallow Anchors
¢;
(b)
Variation (7)--Eq. (7)
62
TRANSMISSION
LINE TOWERS
FOUNDATIONS
or
2 [1
+
m
1
I
I
(D B) ] (D B) Ku tan ¢
+
(7)
1
By using Eq. (7) and the values of Ku and m' given in Fig. 4a, the variation of the breakout factor (Fa) with embedment ratio for shallow circular anchors in sand has been calculated and is given in Fig. 4b. Also shown in the figure is the zone of deep anchor condi tion as recommended by ;'ieyerhof and Adams (7). Clemence
and Veesaert's
Theory
(4)
According to this method, the failure surface in soil is assumed to be a truncated cone (for shalloK anchors) as shoKn in the insert of Fig. S. The net ultimate uplift capacity of a circular anchor in sand can be expressed as
100
,/
50
,/
,/
/~
0/7
30
/
20
/
/
7 Deep
// 400
anchor condition
35°
,.
l
10 4
I
/4-
B= ~
,
:;
-
•.
-0
diameter 68 10
I
5
.
:s
2
a
Embedment
Figure
S.
ratio,
Variation of Fq Kith D/B Anchors--Eq. (9)
D/B
and
¢ For Shallow
MODEL GROUP
ANCHORS
63
IN SAND
(8) where Vs=volume of the truncated cone shown Ko=coefficient of lateral earth pressure
in Fig.
5, and
The value of Ko varies from 0.7 to 1.5, with an average of about 1.0. The lower limit of Ko is for the case in which sand is poured by the raining technique, and the upper limit is for the case where sand is compacted around after the placement of the anchor. It can easily be seen that Vs
= ~[B+Dtan(¢/2)]2DY
Substituting
this
into Eq.
(8) and rearranging
Qo [1+ (D/B)tan(¢/2)]2 (~) 4 B 2 yD
+
4Kotan¢
cos2(q>/2)[}(D/B)
+
(D/B)2tan~¢/2)
J
(9)
Using and average value of Ko=l, the breakout factor variation with ¢ and D/B has been calculated and is shown in Fig. S. In this figure, the embedment ratios at which deep anchor behavior starts have been taken to be the same as defined by ~eyerhof and Adams (7). A comparison of the breakout factors shown in Figs. 3, 4b, and 5 shows the following: 1. For a given soil friction angle (¢) and embedment ratio (D/B), Vesic's theory (8) yields a substantially lower value of Fq than those obtained from the theories of Meyerhof and Adams (7) and Clemence and Veesaert (4). 2. For ¢=30° and 35° with Ko=l, Eq. (9) consistently yields a higher value of breakout factor (for similar D/B) than those obtained by using Eq. (7). For ¢=40°, Eqs. (7) and (9) give practically the same variation of Fq for shallow anchors . .). With ¢=45° and Ko=l, Eq. (9) results in lower values of the breakout factor for D/B ~ about 3.5 than those obta in e d fro m 0'1 eye rho fan dAd am s' the 0 ry [E q . (7) ] . Uplift
Capacity
of Horizontal
Group
Anchors
A review of the existing literature shows that the only theoretical study proposed so far to estimate the ultimate uplift capacity of horizontal group anchors is that of ~Ieyerhof and Adams (7). According to this theory, the net ultimate uplift capacity of shallow circular group anchors can be given as
64
TRANSMISSION
Qo(g)
LINE TOWERS
= Qu(g) -h'g = yD2[L' +L"
FOUNDATIONS
+ (-rr/2)SIB]Kutan
where Qu(a), Qo(o)=gross and net ultimate uplift capacity of anchor gr8up, Kogself-weight of anchors in the group and the cap, Kag=weightbof soil located immediately above the anchor group, L'
S'(m-
L"
S'(n-l)
(11)
1)
(12)
where m and n=number of columns and rows in the plan of the group anchor (Fig. 6), and S'=center-to-center spacing of the anchors
I~S' .. ~~~S'
SI
f .--.--e-- • .I
: S'
e •• el:
I
L"=S
t
. I
(n-l).
I
SI
i. --_ft_.8 __Ci t. !4- L'= 4i S I (m-l)
Figure
6.
Hence, for similar can be expressed as
n(%)
=
Qo ( a) b
mnQ o
(100)
Plan D/B
=
YD2[ L'
+
ratios,
Laboratory
the group
efficiency
(~).
Thus
L" + (0/2)SIB]K
Model
(n)
(13)
u
(mn) [(0/2) SyBD2Kutan¢ Present
Anchor
<100
1S gl\'en by Eq. =
of a Group
tancp + \\' ag(lOO)~lOO
(14)
+ \\']
Tests
A total of 49 small-scale laboratory model tests on single and group anchors were conducted in the laboratory in order to compare (a) the existing theories to the experimentally observed net ultimate uplift capacity of single anchors, and (b) laboratory group efficiency variation (with different ancl10r configurations) with the theory presented
MODEL GROUP ANCHORS
IN SAND
65
by ~!eyerhof and Ada;ns (7), i.e., Eq. (14). A total of 9 model anchors were used in the present study. All anchors had a diameter of 2 in. (50.8 mm) and were made out of steel p 1ate s 1/8 in. (3. 18 mm) th ic k . Each an chor was we 1de d ta a vertical steel shaft having a diameter of 1/2 in. (12.7 mm). The length of each shaft was 18 in. (457.2 mm). Holes were drilled in the top of the anchor shaft for attaching the cap which was required for the group anchor tests. Table 1 shows the sequence of laboratory model tests conducted under this program.
24
7 - No. 6 17 44 49 11 40 30 35 S'/B 2x2 2xl 4 6 6ratio, toto 1. 3x3 3xl lxl Table Embedment 0 Tests configuration D/B 3,4,5,6,7,8 1,1.5,2,3,4,5,6 Sequence spaclng 1,2,3,4,5,6 1,2,3,4,6 1,2,3,5,6 aAnchor f ~!ode1 ln 1,2,4,6 group,
1
The model tests were conducted in a box measuring 5 ft x ) ft x 3 ft (depth) (1.52 m x 1.52 m x 0.915 m). The sides of the box were heavily braced to avoid lateral yielding. The sand used for the model tests was angular and had 100% passing No. 10 U.S. sieve, 71% passing No. 40 U.S. sieve, and 0% passing No. 200 U.S. sieve. The uniformity coefficient and coefficient of gradation were 2.14 and 1.2, respectively. The sand was compacted in the model test box by means of raining to an average unit weight of 98 lb/ft3 (15.41 kN/m3). The triaxial angle of friction at this average unit weight of compaction was 37°. The relative density of compaction (Dr) was 68%. In order to determine if the size of the container used for the model tests had any effect on the ultimate capacity of single and group anchors, a few tests were conducted in a box measuring 6 ft x 6 ft x 3 ft (1.83 m x 1.83 m x 0.915 m). Under similar conditions, the ultimate capacities as obtained from this box were not different than those obtained from the box measuring 5 ft x 5 ft x 3 ft (1.52 m x 1.52 m x 1.52 m). For single anchor tests (Tests 1 to 6 as shown in Table 1), the anchor was placed centrally in the test box and sand was poured in 2-in. (50.8 mm) layers until the desired depth of embedment was reached. After that, a steel cable was attached to the top anchor shaft by means of a hook. The cable passed over two pulleys attached to a steel frame.
66
TRANSMISSION LINE TOWERS FOUNDA TlONS
Step loads ',\'ere appl ied to the load hangcr, and the corre sponding deflections were recorded by a dial gauge until pullout occurred. All tests relating to the ultimate uplift capacity of single anchors were repeated three times, since these were used as the base values to determine the experimental group efficiency. The Jilagnitudes of the experimental Qo reported in the following sections are the average of three trials. For group anchor tests (Tests 7 to 49), a desired number of anchors with proper center-to-center spacing were lightly attached to thin steel strips by means of screws. The group assembly was centrally placed inside the test box, and sand was then poured into the box by raining up to the desired depth. After that, the steel strips were carefully removed from the top of the anchor shafts. A rigid aluminum plate measuring 23 in. x 23 in. (584.2 mm x 584.2 mm) with several holes drilled in it was used as the anchor cap. Once the steel strips were removed, the aluminum cap was carefully placed on the anchor shafts. The anchor shafts and the cap were rigidly connected by scrcws. The reason for attaching the anchor shafts to thin stccl strips first was to assure proper sand compaction as much as possible and still maintain proper center-to-center spacing. A steel cable was attached to the top of the pile cap. Other loading procedures were similar to those used for single anchor tests described above. A schematic diagram of the laboratory test arrangement is shown in Fig. 7. For all group anchor tests, the failure surface did extend to the surface signifying shallow anchor conditions. Laboratory Ultimate
Test Uplift
Results Capacity
of Single
Anchors
During the laboratory tests, the net load on the anchor increased with the vertical movement of the anchor, and failure occurred by sudden pullout of the anchor. The vertical anchor displacement at which the net ultimate load was reached increased with the embedment ratio, D/B, varying from about 4 mrn to 8 mm, signifying that the failure load occurred at a displacement of 8-16% of the anchor diameter. The net ultimate load, Qo, as determined from the laboratory experiments is shown in Fig. 8a. In order to compare the present experimental results with various existing theories, the experimental breakout factors at various embedment ratios have been calculated and are shown in Fig. 8b. Along with this, Fig. 8b also shows the theoretical plots as obtained from the theories of Vesic (8), ~Jeyerhof and Adams (7) ~ and Clemence and Veeseart (4). From this, the following conclusions can be drawn. 1. The experimcntal value of the breakout factor increases with the embedment ratio and remains practically
MODEL GROUP ANCHORS IN SAND
Cable Pulley
Dial
.
rl
,
r
.. •••••••••
Z 3 .-~ Q) / Anchors nchor; Group
I
'"
~
: .~ . .
.
.
: . -:'
j
-
..
:..
-L-L
..
~
,
'..•...... ......
~
••••
O'
•••
~
J
.: . ' : .. ... . '--1" ., ..... Sand
;.. : .'
,/ / I ,,/ / /.---e--f
,
::...
gauge
'
~V) 45 / ITheoretical Ir410 73 6I q Diagram F I4(a) 0 30 20 I-J 5II8. D/B Comparison Schematic of Uplift Model Test and Arrangement Experimental Net Ultimate / /I of Capacity of Single rl::1 Figure I 50 (b) I6I,;',(b)8\_T"hD~"""'T Experiment
Breakout
Factor
For
I Single
Anchor
67
TRANSMISSION LINE TOWERS FOUNDATIONS
68
constant beyond D/n~6.5. D/B=(D/Bcr=6 as predicted
Eq.
This is fairly close to a value of by Meyerhof and Adams (7) and also
(1).
2. The present experimental values of F for D/B<6 are close to wh~t has been predicted by Meyer]lo~ and Adams (7). The theoretical variation of Fa as given by Vesic (8) is substantially lower than the experimental values. 3. The theory of Clemence and Veesaert (4) gives slightly higher values of Fq than the experimental results for shallow anchor range. However, for deep anchor condition, the agreemen~ of the magnitude of the breakout factor is good. Ultimate
Uplift
Capacity
of Group
Anchors
The experimental ultimate uplift capacity of group anchors listed in Table I (Tests 7 to 49) were determined from the load-displacement plots. As in the case of single anchors, failure occurred by sudden pullout. Using the conventional definition of the group efficiency as given in Eq. (13), the experimental variations of n vs. St/B have been determined and are shown in Figs. 9 and 10 for D/B=4 and 6, respectively.
---a ---
-- -.---
Trleory; Eq. (14) 2xl; --Expt.
• ---
3xl;
---T ---
Expt. 2xl; Expt.
100
80
I
63 42 5
40
20
60
Figure
r
3xl 2x)
9. Comparison of Theoretical and Experimental Variation of Group Efficiency--D/B=4
MODEL GROUP ANCHORS
___ •
Theory; Eq. (14) __ -..
3x1; Expt.
---A----
2xl; Expt.
69
IN SAND
2x2;
Expt.
-..--3x3; Expt.
100
80
....,
u
.~
2xl
60
",
,,/
40
/
/
/'"
/ ~
2x2
20 o
2
4
6
8
10
12
S' /B
Figure
10. Comparison of Theoretical and Experimental Variation of Group Efficiency--D/B=6
In order to compare the present experimental results and ~!eyerhof and Adams' (7) group efficiency theory, Eq. (13) has been used to calculate the variation of ~ vs. S'/B. These values are also shown in Figs. 9 and 10. A comparison between the theory and experimental results shows the following: 1. According to the theoretical prediction. for a given soil type, compaction, and embedment ratio, the uplift efficiency of a given group anchor increases practically in a linear manner with S'/B to reach 100%. The present experimental results show a generally similar trend. 2. For a given group magnitude of experimental predicted by Eq. (14). J.
D/B=4
configuration. D/B, and S'/B, the ~ varies substantially from that
According to the present tests, group anchors with reached an efficiency of 100% at S'/B~4.S to 5.5
TRANSMISSION LINE TOWERS FOUNDA nONS
70
(i.e., S'/B about 1.25D/B). However, the theoretical value of S'/B for ~=100% is approximately 3 (i.e., 0.75D/B). 4. For group anchors with D/B=6, the experimental efficlency cf 100% was reached at S'/B~6 for group cOJlfigurations of 2x1 and 3x1. However for group configurations of 2x2 and 3x3, a value of n~about 90% was reached at S'/B:6. Although no experiments were conducted beyond S'/B=6, the projection of ~ vs. S'/B plots for these configurations show that the efficiency might have reached 100% at S'/B~7 to for these 7.5. The theoretical values of S'/B for n=100% cases varies between 3.5 to 4.5 (i.e., S'/B~0.75D/B). 5. In general. for a given D/B and SI/B, the group efficlency decreases with the increase of the number of anchors in the group. 6. clency
For a given SI/B and group configurations, the effidecreases with the increase of embedment ratio.
Conclusions The results of the laboratory model tests for ultimate uplift capacity of shallow circular single and group anchors embedded in medium dense sand have been presented. A maximum of 9 anchors in a group was used for the present tests. Based on the present study, the following conclusjons can be drawn: 1. The ultimate uplift capacity of single shallow Clrcular anchors in medium sand agrees well with those predicted by the theory of jljeyerhof and Adams (7). The magnitudes of Qo predicted by Vesicls theo~y and Clemence and Veesaert's theory are too low and too large, respectively. 2. The efficiency of shallow circular group anchors in sand depends on several factors such as the degree of compaction of sand, embedment ratio, number of anchors in the group, group configuration, center-to-center spacing of anchors, etc. Foy medium dense sands as used in these tests, the experimental group efficiency reaches 100% at S'/B~1.25 D/B. 3. The group efficiency of an anchor group follows a generally similar trend as predicted by the theory of Meyerhof and Adams (7). However, the magnitude of ~ varies widely from those predicted by the theory. The experimental values of S'/B at which ~=100% is obtained is about 1.25 to 1.5 times the value predicted by theory. 4. The group efficiency of horizontal anchor groups decreases with the increase of anchors in the group, centerto-center anchor spacing, and embedment ratio (D/B).
MODEL GROUP ANCHORS
IN SAND
71
References 1.
2.
3.
4.
S.
Adams, J.1., and Hayes, K., "The Uplift Capacity of Shallow Foundations," Ontario Hydro Research Quarterly, Vol. 19, No.1, 1967, pp. 1-12. Baker, W. H., and Kondner, R. L., "Pullout Load Capacity of a Circular Earth Anchor Buried in Sand," Highway Research Record No. 108, National Academy of Sciences, 1967, pp. 1-10. Balla, A., "The Resistance of Breakout of ~Iushroom Foundations for Pylons," Proceedings, V International Conference on Soil Mechanics and Foundation Engineering, Paris, Vol. 1, 1961, pp. 569-576. Clemence, S.P .. , and Veesaert, C.J., "Dynamic Pullout Resistance of Anchors in Sand," Proceedings, International Symposium on Soil-Structure Interaction, Roorkee, India, 1977, pp. 389-397. Das, B.M., and Jones, A.D., "Uplift Capacity of Rectangu 1a r Founda tions in Sand," .I.T_~!l.2P0rta t ion Re se arch Record No. 884, National Academy of Sciences, 1982, pp. S4 - S8 .
6.
7.
8.
--_.
Esquivel-Diaz, R.F., "Pullout Resistance of Deeply Buried Anchors in Sand," M.S. Thesis, Duke University, Durham, N.C., 1967. Meyerhof, G.G., and Adams, J.1., "The Ultimate Uplift Capacity of Foundations," Canadian Geotechnical Journal, Vol. S, No.4, 1968, pp. 224-244. Vesic, A.S., "Breakout Resistance of Objects Embedded in Ocean Bottom," Journal of the Soil :Vlechanics and Foundations Division, ASCE, Vol. 97, No. SM9, 1971, pp. 11831205.
Acknowledgements In-depth studies relating to the subject described in this report, as well as determination of the dynamic uplift capacity of anchors, are presently being pursued under ~ational Science Foundation Grant No. RII8604l32. This support is greatly appreciated.
HELIX ~~CHOR FOUNDATIONS--TWO Albert M. Weikartl,
CASE HISTORIES
M ASCE, and Samuel P. Clemence2, M ASCE
ABSTRACT Two case histories are presented which describe the site conditions, foundation design, construction, and performance of transmission towers supported on helix anchor foundations. Both sites are located in low lying marshes in Central New York. Access for site exploration and construction was limited. The foundation design was based on minimal information of soil properties and was modified in the field due to installation problems. A comparison of the foundation capacities based on an estimate from installation torque is made with capacities based on geotechnical parameters of the soil. Construction procedures in difficult terrain and resolution of problems encountered during construction are described. Introduction The construction of structural foundations in remote, low lying marshes presents a challenge to the geotechnical engineer. The case histories described in this paper describe the use of helix anchors as a successful foundation system ror electric transmission towers in the Rattlesnake Gulch and Bear Swamp sites in Central New York. Both of these sites presented problems in terms of poor foundation material and remote location with limited access. The construction of conventional foundation systems was precluded due to the high water table and difficulty of access for equipment and materials. Helix anchor foundations provided a viable alternative which could be installed under difficult conditions with moderate equipment support in a short amount of time. The helix anchors were used to support four towers at the Rattlesnake Gulch site and four towers at the Bear Swamp site. (The foundations were installed in the summer of 1975 and winter of 1975-76 and have performed successfully for the past twelve years.) The site conditions, foundation design, and construction will be discussed for each site. Rattlesnake
Gulch Site
During the summer of 1974, Niagara Mohawk Power Corporation's line department began an effort to repair the deteriorated foundations on the Teall-Oneida #2 and #5 11SkV double circuit transmission line.
IStructural York.
Engineer, Niagara Mohawk Power Corporation,
2professor and Chairman, Civil Engineering sity, Syracuse, New York.
72
Department,
Syracuse, New
Syracuse Univer-
HELIX ANCHOR
FOUNDATIONS
73
The line was constructed in 1913. According to descriptions of older residents of the area, the primary construction equipment was horses and stone boats. The area known as "Rattlesnake Gulch" is a swamp in which the water table has risen over the years. It is reported that during the 1940's the line blew over and the towers were pulled back up and guyed. According to the story, a horse became mired in the mud. Attempts were made to pull it out but failed and the horse had to be destroyed. During the repair program in 1974, the skeleton was uncovered confirming the story and the hazardous nature or the site. The line is composed of steel flex (two dimensional) towers on steel grillage foundations. Due to the soft soils, rising water table, and marginal design, the foundations were rising on one side and/or settling on the other side resulting in towers leaning in response to the prevailing winds. It had been recognized that replacement grillage type roundations would not be appropriate. Assuming the need for heavy equipment to install a deep foundation such as piles or drilled piers, construction of an access road was initiated. In 1974 before beginning a tower repair program, the line department initiated construction of a routine gravel road to provide access for construction vehicles. As the road progressed into the softer areas, more gravel was required. When the gravel requirement became excessive, a geotextile, Mirafi 140, a new product at the time, was placed on the ground and gravel was placed on it. The "magic carpet" floated the road on top of the swamp eliminating the need for enough gravel to build a road up from the bottom of the swamp. As the road construction progressed onto the deeper part of the swamp, there was insufficient bearing capacity to support the road, and it sank overnight leaving a long narrow pond. A road was initiated along the right of way from the opposite side of the swamp but also ended in a pond. The shallow bearing capacity railures of the road created mud waves and displaced one tower several feet transversely off the centerline of the line. Since the access roads would probably not withstand repetitive heavy traffic and vibration and since one of the leaning towers needing foundation repair was isolated between the terminations of the two roads, it was apparent that the usual piles or drilled piers could not be used. Something new and difrerent would be required. Foundation
Design
The site is located 10 miles (16 km) northeast of Syracuse, New York in the Erie-Ontario plain region of ~ew York State. The soil profile was developed based on three test borings made along the transmission line route. The soil profile consists of three to four foot (0.9-1.2 m) layer of gravel fill underlain by six to eight feet (1.8-2.4 m) of very soft peat, muck, and marl. This organic layer is underlain by 20 to 24 feet (6.1-7.3 m) of soft organic silts. These sort soils are underlain at depths from 30 to 38 feet (9.2-11.6 m) by compact to dense interbedded silts and fine sand. The borings terminated in the dense silt and sand at a depth of 50 feet (15.3 m). The area is very poorly drained (swampy) resulting in a water table at or very near the ground surface. The soils were saturated throughout
74
TRANSMISSION
LINE TOWERS
FOUNDATIONS
the profile. Figure 1 gives the profile with N values from standard penetration tests taken during the subsurface investigation. The bottom design are as
transmission towers were double circuit 45 foot (13.7 m) (at the arm) steel flex towers supporting two 115 kV power lines. The loads based on high wind and heavy ice conditions for each leg follows: Horizontal Load: Vertical Load:
5,000 Ibs (22.3 kN) Bearing: 55,000 Ibs (244.8 kN) Uplift: 48,500 Ibs (215.8 kN)
A deep foundation system was required due to thick zone of sort soils overlying the compact, dense sand layer at 30 to 38 feet (9.2-11.6 m) below the surface. Due to the depth of soft compressible soils at the site, a shallow type foundation was ruled out. Driven piles were considered, but there was a concern for safety with the heavy pounding or vibrating equipment on the road. There was already evidence that the vegetation mat was failing where subjected to repetitive traffic. A local contractor did submit a price, but it was high and did not provide a solution to correct the isolated structure. The concept of the combined tension compression helical anchor foundation was selected. Helix anchors had been used widely as tension anchors primarily for guy applications and in limited applications for bearing type foundations. To protect their own R&D work, the vendors were willing to recommend an anchor for the application but would not
8-1
o
-.W W
-97/ (GRAVEL 2 -:::'--
10
2
--
I
L.1..
I.-
---
=..::. '=-PEAT
20
CL
o W
30
40
.
~~. DENSE' FI NE . 19'SAND AND SiLT·.··
2 i·.. . ·. :' . '.: ~I
Figure 1.
•-
~
.',
..... :
:
••
.
••• "
.
I
, ••..• , I ••. I
,
,. -
<
..
Subsurface Profile for Rattlesnake Note 1 ft = 0,305 m
. .., Gulch
. .
HELIX ANCHOR
75
FOUNDATIONS
provide an analytical basis for the design other than some empirical chor selection charts.
an-
Since access was limited and movement of heavy construction equipment into the area was not feasible, a helical anchor foundation which could
1SOFT SOILS
/ .•
..-
,
~, ...
'.
,,-'-.DENSE SOIL
3 HELIX ANCHOR (01 AM. 11.3,10 and 8 inches)
Figure 2.
Typical Anchor Configuration Note 1 inch = 2.54 cm
in Soil Profile
bs)
TRANSMISSION
76
LINE TOWERS
FOUNDATIONS
be installed using relatively light torque installation equipment was selected. The helix anchor foundation system is shown in Figure 2. To sustain the compressive and tensile loads from the tower legs through the soft soils, an eight inch (20.3 em) diameter, Schedule 40 steel pipe was selected for each tower leg. The bottom of the pipe was attached to a triple helix anchor with plate diameters of 11.3 in, 10.0 in, and 8.0 in (28.7 em, 25.4 em, 20.3 em). The specifications from the anchor company required that the anchors must be installed in the dense soil layer at least three to four reet above the top helix (0.9-1.2 m) with an average installation torque or 5,000 ft-lbs (6.78 kN-m) to develop capacity ror the anticipated design loads. During construction the recommended penetration depths could not be achieved due to the compact nature of the dense sand and silt layer. The average depth of penetration ror the top helix for each tower leg was only two feet (0.6 m) into the dense sand; the anchors refused at approximately 8000 ft-lb (10,850 N.m) torque. The maximum torque criteria was easily achieved; however, the required minimum depth or penetration was not met. This limited penetration raised questions as to the ultimate uplift capacity or the anchor. In order to estimate the anchor capacity based on geotechnical parameters, an analysis was performed using uplift design procedures recommended by Mitsch and Clemence (2) and Goin (1). The estimated capacities along with the estimate based on installation torque are shown in Table 1. The calculated values neglect any skin friction which may develop on the sides of the eight inch (20.3 em) diameter pipe in the soft silts and organic soil above the dense soil layer. The calculated values are in fair agreement with those estimated from installation torque. Table 2 shows anchor uplift capacity relationships. Table 1.
Comparison Site
of Anchor Uplift Capacity for Rattlesnake
Gulch
---
qA.B. Chance u ~Mitsch k Prediction qu and 1.2 ~ (lbs) (lbs) (1) 26 Torque 80,000 HID y' (pet) 69,250 Clemence (2) 38I 52,400 60 151 0.5 I &,chor Uplift 34 Capacity 9 Soil PropertiesI
I I
60231 ,
~ote 1 ft = 0.305 m, 1 Ib/ft3
= 0.157 ~~/m3,
1 Ib = -4.5 N
The field performance of the towers confirms that sufficient uplift and bearing capacity has been developed by the anchor foundation. The towers have withstood severe wind and ice conditions with no movement or damaging settlement. Construction The foundations
of the three towers accessible
via the access road
HELIX ANCHOR
FOUNDA TrONS
77
~ere reconstructed first. An 8400 ft-lb (11,400 ~.m) drive head was fitted to a boom mounted at the center of the flat bed truck to minimize ~eight and eccentricity. All tools and materials were loaded on separate light trucks. The fourth structure was accessed with a tracked machine on packed snow the following winter. The installation was routine at all four sites with only minor problems. One anchor struck the old grillage foundation but deflected slightly and penetrated normally. Clearance for the anchors to miss the grillages had been allowed in the design, but this foundation had apparently been distorted years earlier. Another anchor was deflected by a boulder near the surface, but the pipe column was successfully pulled back within the allowed tolerances with a winch on a pick-up truck. A crane was used to align the displaced tower on its new foundation. The other towers were attached to the new foundations without being replumbed to avoid the risk associated with utilizing a crane. Table 2.
Relationships
1.
for Anchor Uplift Capacity
Mitsch and Clemence
(2) - Deep Anchors:
+ -2 D a'y iT
2.
A.B. Chance
3.
Installation
'(
k
U
Torque in ft-kips)
anchor uplift capacity
friction angle of soil area of top, middle, and bottom helix
N
uplift capacity
. qu
factor ror sands
depth to top, middle, and bottom helix
"1,2,3
Note:
u
effective unit weight of soil lateral earth pressure in uplift
A 1,_, ? 3
p
1
Torque:
ultimate
'11
D
3
(1) - Deep Anchors:
Q u (kips) = 8(Installation Qu , y
" 2 -" .2) k tano
a
average helix diameter
s
perimeter 1 kip
or anchor shaft
4.45 kN, 1 ft
0.305
m
Bear Swamp Site Monday evening, June 23, 1975, a freak gust of wind associated with a thunderstorm, caused a collapse of four square base lattice towers on
TRANSMISSION
78
LINE TOWERS
FOUNDA TrONS
the Clay-GE #14 double circuit line. A fifth structure suffered a buckled leg post member but did not fail. It was fortunate that the transmission system had grown up in a manner that provided several alternatives to the failed line because the failed towers were situated near the center of the Bear Swamp located north of Syracuse, New York, and it is unknown how restoration crews would have repaired the damage at the time. Due to fill placed as the northern suburbs have grown up, the water table in the Bear Swamp has risen significantly since the line was built around 1920. Fortunately, the towers fell just short of the parallel railroad tracks so the trains could pass, slowly, on the main line to Watertown and Massena, New York. Due to the site conditions and based on experience from Rattlesnake Gulch, the applicability of combination tension compression helix anchor foundations was recognized. Use of equipment parked on railroad cars to install the foundations and erect the towers was considered, but the railroad schedule would have severely limited the work periods. A local contractor was contacted to consider his fleet of low bearing pressure tracked equipment. One of the smaller machines served as a platform for the soil investigation, but it was recognized that the vegetation mat would not support a larger machine. Foundation
Design
The site is located approximately
l.J... •
>
~
w tI0 w
o
..
30 50 40
....
~
eight miles
(13
kID)
..
70 60
Figure 3.
Subsurface Profile for Bear Swamp Note I ft = 0.305 m
north of Syra-
8-2 ---
HELIX ANCHOR
FOUNDATIONS
79
cuse, New York in the Erie-Ontario plain region. Five borings were made along the transmission line right of way. A typical soil profile is shown in Figure 3. The surficial layer consists of five to seven feet (2.4-3.0 m) of very soft organic muck and peat. The organic layer is underlain by eight to ten feet (3.4-4.8 m) of cospact fine sand. The sand is underlain by a thick layer of interbedded sort varved silts and sand which range from 25 to 45 feet (7.6-13.7 m) in thickness. The varved silt is underlain by a dense fine sand and gravel layer. Borings were terminated at depths of 52 to 75 feet (15.9-22.9 m). The water table was at the ground surface in all the borings. Figure 3 also includes the standard penetration test values (N) taken during construction. The replacement transmission towers are double circuit 45 foot (13.7 m) (at the lower arm) steel flex towers to support two 115 kV power lines. The flex towers were selected for their light weight anticipating erection by helicopter. The design loads and foundation design were the same as the Rattlesnake Gulch site. The anchors were intended to extend into the dense fine sand and gravel. Installation, however, of the anchor was also difficult in the compact sand layer with the top helixes penetrating an average of one foot (0.3 m) into the compact sand. A comparison of uplift capacities based on analyses by Mitsch and Clemence (2) and Goin (1) is showu in Table 3. The uplift capacities are based on uplift resistance from the anchor helixes only neglecting skin friction along the eight inch (20.3 em) diameter pipe in the soft soils overlying the dense sand. All of the calculated capacities are well in excess of the design loads. These to~ers have performed with no problems since installation under ice and ~ind conditions. Table 3.
¢
Comparison
of Anchor Uplift Capacity -
u q qu )l HID (pc f) Clemence (2) )l k I y'(lbs)
I
Mitsch and
for 3ear Swamp Site
Prediction Anchor Uplift Capacity (1) (lbs) (lbs)Torque I
I
A.B. Chance
peat
I 55,110
Compact Fine
sand! )lote 1 ft = 0.305 m,
86,000
100,000
6 31 ~~ 8 Ib = 4.5 ~
Construction The decision was made to wait for winter when it was possible to pack down the snow with a small tracked machine and thus build up a frozen road which was reinforced with slab wood as well. A torque head was fitted to the boom on a large flexible track machine and a hydraulic pressure gauge calibrated to the torque rating of the head was installed to provide direct torque readings. Plans were made to install the foundations and erect the flex towers from the tracked rig. Although the foundation anchors were intended to be embedded with the top helix at least two feet (0.6 m) in the sand layer 50 to 70 feet (15-
TRANSMISSION
80
LINE TOWERS
FOUNDATIONS
20 m) down, the anchors refused when the top helix was one foot (0.3 m) into the fine sand at 10 to 12 feet down. Excessive torque only resulted in rotation with no penetration. In an effort to confirm the soil boring data, a double ten inch tension anchor was installed nearby. It drove to more than 30 feet (9 m) at 2000 ft-lb (2730 K.m) of torque. The adhesion and cohesion nearly prevented withdrawal of the installing wrench from that depth. The foundation anchor had performed properly, however, once the tip of the column formed a sand plug, it could not displace enough sand to penetrate any deeper. Since the high (rerusal) torque was due to the pipe column and not the helix anchor, there was concern that the anchors could rail in uplift. Also, there was concern that in bearing, the anchors could punch through the compact fine sand layer into the soft varved silt below. Therefore, to supplement the foundation anchors, smaller tension compression anchors were fabricated from double ten inch (25.4 em) guy anchors and three inch (7.6 em) Schedule 80 pipe. They were installed into the sand layer and framed into the tower base. Warmer weather arrived as the supplemental anchors were being installed. The flexible track machine destroyed the road on the way back out. The following summer helicopters were employed to erect the towers and lift the wires back up to the towers from the swamp where they had been for a year. Summarv and Conclusions Helix anchor foundations provided a viable alternative deep roundation system at sites where limited access and struction conditions were encountered. The installation field provide userul inrormation ror future use of these
to a standard dirficult conproblems in the roundations.
The results rrom these two sites indicate that pipe columns will not advance into a compact fine sand layer. The designer should be aware of this installation limitation when using helix anchor-pipe column roundations. A comparison or predicted uplift capacities based on installation torque and an analysis based on geotechnical parameters indicate that both methods are useful in estimating uplirt capacity. The tower roundations have perrormed successfully ror a significant length or time during periods of ice and wind conditions. References 1.
Goin, J.L., "Design Examples of Helical Anchors," Foundations Tension, Seminar Notes, Kansas City, MO, October 2, 1986.
in
2.
Mitsch, M.P., and Clemence, S.P., "The Vplift Capacity of Helix Anchors in Sand," Uplift Behavior of Anchor Foundations in Soil, ASCE, October 1985, pp. 26-47.
HIGH CAPACITY MULTI-HELIX SCRE'WAOCHORS FOR TRANSMISSICN LTh1EFOUNDATICNS Thomas E. Rodgers, Jr.* Abstract Three case history summaries are presented which discuss the siting, design, and construction of towers supported on multi-helix screw anchor foundations in the Virginia Power service area. Access for ~xploration and construction was difficult. The procedure used for construction and the problems encountered during construction are described. Introduction In the past, Virginia Power has been called on to rebuild transmission lines in eastern Virginia and northeastern North Carolina as part of a program to upgrade service to areas which are ~xperiencing industrial and residential gr~vth. Each project authorization called for the replacement of existing wood H-frame 115 kV lines with 500 kV or 230 kV lines. Each project appeared to be relatively routine. However, preliminary engineering reviews of aerial photos and geodetic maps revealed a basic fact which would greatly change the engineering approach to a portion of each line. The lines lay in the geomorphic Eastern Coastal Plain Province which is characterized by a gently sloping flat regional surface with wide flat flood plains in the form of tidal marshes or swamps. The ~xisting line routes dictated the crossing points, and it was obvious that major access and environmental problems would be encountered. Engineers were faced \vith the task of developing a combination of structures and foundations which could be constructed by enviropmentally compatible means. These circumstances eventually lead to ~~e use of power installed multi-helix screw anchor foundations. Multi-helix screw anchors are often used to support a variety of high voltage transmission line structures such as free-standing and guyed lattice towers, guyed pole and guyed H-frame structures. Structures of this type generate very large base reactions when subjected to wind and ice loadings and, therefore, require foundations and anchors capable of resisting enormous compressive and uplift forces. The use of high capacity multi-helix scr~N anchors for these *Civil Engineering Manager (T&D), Virginia Power, P. o. Box 26666, 7th & Cary Street, Richmond, Virginia 23261.
81
TRANSMISSION
82
LINE TOWERS
FOUNDATIONS
applications is especially attractive when the transmission line right-of-way 1) is located in areas where near-surface soil conditions are inadequate to accommodate heavy construction equipment, 2) is in remote areas where mobilization of such heavy equip.~nt is inconvenient and costly, and/or 3) is inaccessible to oversized equip.~nt due to undesirable topography and/or dense vegetation. Equipment typically used to install screw anchors consists of a mechanical digger or earth auger for positioning and advancing the screw anchors. Equipment of this type is relatively light, when compared to large cranes, pile drivers and concrete trucks used to install conventional foundations and, therefore, has minimum impect on sensitive environments such as wildlife refuges of coastal marshes. Tne screw anchor installation process is typically a one step operation eliminating the need for temporary casing, concreting, and/or select compacted backfill processes. Another advantage of screw anchors is their ability to provide the full ultimate capacity immediately after installation, which could result in a substantial savings in the total transmission line construction time and cost. Description
,. j!
~
II ,.
Ii
=====
• Ii
= = Ii
1 Ii
o
Fig. 1 Typical ~mlti-Helix ScrBv Anchor
of a Screw Anchor System
A typical multi-helix screw anchor system (Fig. 1) could be composed of 1) a lead section equipped with two to four helices spaced as close as 30 inches (760 rom), of varying or identical diameters ranging from 8 to IS in. (200 to 380 rom), 2) anchor extension sections wiL~ one to four helices, 3) a series of 1.5 in. to 10 in. (40 to 250 rom) solid connecting rods or extension pipes having a square or circular cross-sectional configuration, and 4) a guy adaptor or base plate. All components are generally forged from high strength corrosion resist~Dt steel. Lead sections and extensions generally come in 3.5, 5 , 7 and 10ft. (1, 1.5 , 2 , and 3 m) lengths. Helices of varying sizes welded on a lead section decrease in diillnetertowards the tip or the section. Some commercially available anchors are described in the "Encyclopedia Anchoring" [1] and the "PCM1er Installed Screw Anchor Handbook" [2].
0:
MULTI-HELIX SCREW ANCHORS
83
O.5E HISTORIE'S Site Description and Geology The three transmission line segrrents to be examined in the Coastal Plains Province are: 13 mi. (21 km) on the Suffolk-Yadkin 500 kV line, engineered in 1968 and built in 1969-70; 1.25 mi. (2 kIn) on the Lane..xa-Shackleford 230 kV line, engineered in 1974, built in 1976-77; and 3 mi. (5 kIn) on the Earleys-Trowbridge 230 kV li.1'1e,engineered in 1977, built in 1978. The Suffolk-Yadkin 500 kV transmission line crossed the northern end of one of the geological wonders of the area, The Great Disrral Swamp. Originally, the swamp spread over approximately twenty-two hundred square miles of dense, partially inundated forest 1~1'1d in southeastern Virginia and northeastern North Carolina. Thousandsof acres of the swampland have been cleared and drained for cultivation. There are many miles of dry forest around the edges of the swamp. Today the swampproper contains between seven hundred fifty and one thousand square miles and is about 40 mi. (64 km) long, running north and south, and 15 mi. (24 kIn)wide east and west. (Fig. 2). The water in this swampyreservoir is trapped by the land escarp:rent on the west, a sedirrentary sea bottom underlying the swamp which is linpervious to water, by rows of sand dunes on the seaward side, and densely entangled undergrowth in and about the swamp. The floor of the swamp composed of bark, woodand juniper leaves is a quagmire locally ~~owTI as "scurf". This huge sponge remains water soaked and is so soft that horses and mules find it difficult to walk on, and it trembles under manI s feet. Soil borings along the line reveal that the Dismal Swamp peat is highly variable in L~ickness, 2 to 20 ft. (0.61 to 6.1 m), because it was deposited on an irregular topography. The peat consists of soft, spongeliJ<.e masses of decaying leaves, twigs, st1.:IIrq?s, logs and other plant debris. It is highly compressible, is sheared easily, and is accompanied by a high water table. Draining the swampis unfeasible because of the potential for fire damageand atrrospheric oxidation of the peat. The Lanexa-Shackleford 230 kV transmission line, (Fig. 3) crosses approximately 1.25 mi. (2 km) of tidal marsh from G'1ehigh ground on the south edge to the Pamt1f\.key River. The Eltham i-1arshis a tranquil tidal marsh located on the Pamunkey River, immediately above the confluence of the Mattaponi and L~e PamunkeyRivers at West Point, Virginia. The marsh cover is predominantly marsh grasses with sparse scatterings of scrub trees or shrubs. A network of meandering canals criss-cross the area. This type of marsh is considered to be very important with respect to the environment and, therefore, was given careful attention by Virginia Power. Soil borings taken in the marsh
84
TRANSMISSION
LINE TOWERS FOUNDATIONS
....
..
..,.~ "J
'
~
f--
.~ "~ f.-. _ i. '" ·'.f
I
Fi
g.
2 ;.,'
DISMAL
SWAMP
-,'c'· _/'0
!;
/:::-;;:
MULTI-HELIX
FIG.
Fig.
4.
3.
SCREW ANCHORS
ELTHAMMARSH
ROANOKE SWAMP
85
86
TRANSMISSION
LINE TOWERS
FOUNDATIONS
along the transmission line show a layer of organic silt ranging from 25 to 75 ft. (7.6 to 33 m), with no strength, underlain by a green fine silty clayey sand marine deposit. In North Carolina, the Earleys-Trowbridge 230 kV transmission line, which occupies a new right-of-way, crossed approximately 3 mi. (4.8 krn) of a densely forested cypress swamp in the Roanoke River flood plain (Fig. 4). The ruggedness and the vast area covered by this swamp created major access problems. Access to the right-of-way was attainable at three locations; 1) from the high ground on the south, near Trowbridge Substation, 2) from Broad Creek approximately one-third the way into the swamp, and 3) from the Roanoke River on the north end. Soil borings taken at these access points, along the transmission line, showed peat deposits varying in depth from 20 to 25 ft. (6 to 7.5 m) which are underlain by loose to firm fine sands. Tnese soils extend down to a depth where the marine deposit of dense silty sand or stiff silt and clay with shell fragments are encountered. Engineering Virginia Power chose to build the section across the Dismal SWamp on a structure adaptable to helicopter erection. Tne tower chosen was a single circuit 500 kV guyed Y aluminum lattice structure with a ruling span of 1200 ft. (366 m). The tower was designed to withstand a hurricane wind velocity of 105 mph (169 krnph). The tower weighed 4300 lb. (1952 kg) and waS to be erected in one piece by helicopter. The guyed tower required a base foundation which would withstand a vertical compression load of 112,000 lb. (50848 kg) and lateral loads in the transverse ~~d longitudinal direction of 2500 and 4500 lb. (1135 and 2043 kg), respectively. The four guys restraining the tower were to be designed to t~~e the remainder of the transverse and longitudinal loads. A longitudinal profile was developed from the fifteen soil borings (Fig. 5) taken through the swamp. The profile showed varying thicknesses of peat and organic clay with some silt, over sands and silty clay of varying densities. This material was all above what they call "the sedimentary sea bottom material" which is an overconsolidated silty fine sand with shell fragments of the Miocene Age. The depth to this Miocene material varied from less than 10 ft. (3 m) at both edges of the swamp to more than 70 ft. (21 m) along the profile. The tower base foundations, as designed, are assemblies made up of three multi-helix screw anchors (Fig. 6), on a 100 batter away from the center with a Y-shaped grillage that rests on top of the three screw ~~chors (Fig. 7). Each screw anchor consists of a 10 ft. (3 m) lead section 3.5 in. (76 rom) in diameter, with three helices of varying diameters, 10, 11.3, and 13.5 in (254, 287, fu~d 342 rom) on 36 in. (915 rom) spacL~gs; a 10 ft. (3 m) extension section with four 15 10 ft. (3 m) lengths of 8 in. (380 rom) diameter helices; and as ~~y in. (203 rrm) pipe as it took to get the required depth. Each anchor
MULTI-HELIX SCREW ANCHORS
was to te installed (13560 N·m) ~ muchas possible.
87
to 9000 Th.ft (12195 N'm) rnin.imurn and 10,000 Th·ft torque so as to penetrate the Miocenematerial as
The guy tension anchor, as designed, was a multi-helix screw anchor utilizing a 1. 5 in. (38 rrm)square steel bar for the rod section. Each anchor consisted of a 10 ft. (3 m) lead section with four helices 8, 10, 11. 3, and 13.5 in. (203, 254, 287, and 342 rrm)on 36 in. (915 rrm) spacing and a 10 ft. (3 m) extension section. fach anchor was to be installed approximately 30° off the vertical, to 4500 Th.ft (6100 ~m) minimumand 5300 Th.ft (7186 N.m)IM.Ximum torque. Installed, the guy anchor should give an ultimate holding p:1Ner of 70,000 Th (31780 kg) .
After the installation of the first complete tower foundation system, a static load test of the tower base foundation and a guy anchor was rrade under the supervision of A. B. ChanceCompanyand observed by Virginia Power. The static load test of the tower base foundation unit was to 125,000 lb. (56750 kg). Loading increments of 20,000 to 100,000 lb. (9080 to 45400 kg) and then in 5,000 lb. (2270 kg) increments to 125,000 lb. (56750 kg) were applied and settlement readings were rrade at each of the load incr~~ts. Under the IM.Ximumcompression loading conditions, the foundation settled 0.25 in. (6 rrm); upon release of the load, the unit recovered to its original elevation. The first test of the guy anchor, which was being tested to 60,000 lb. (27240 kg) failed at 50,000 lb. (22700 kg) with a steady creep. The 40 ft. (12 m) anchor was placed at a 60 degree angle giving it a 34 ft. (10.4 m) vertical orientation from the ground surface. A second anchor was installed with the helix section of the design illlchor plus a 10 ft. (3 m) extension with two 13 in. (330 rom)helices. It was installed to 4500 lb·ft (6100 N.m) torque at 30 ft. (9 m) along the rod or a 25 ft. (8 m) vertical measur~rnent below the surface. This anchor held up to 69,000 lb. (31326 kg) and failed under a steady creep at 70,000 lb. (31780 kg) . The guy tension anchors installed were changed from the initial design to the guy multi-helix anchor consisting of a 10 ft. (3 m) section with four helices, 8.0, 10.0, 11.3 and 13.5 in. (203, 254, 287 and 342 rrm) in diameter on 36 in. (915 rrm)spacing, a 10 (3 m) foot ~xtension section with two 13.5 in. (342 rom) diameter helices and extension rods. The anchors were to be augered until a minimumof 4500 lb·ft (6100 N~m)torque was reached. Each guy anchor assembly installed was field tested to 25,000 lb. (1135 kg) tension. The purpose was to test the anchor to working load, set the anchor, and check its alignment. For the 230 kV transmission line across the Eltr2rn Marsh, six 155 ft. (47 m) self-supporting double circuit lattice towers were to be used. The maximumfoundation reactions are 100,000 lb. (45400 kg) tension, 123,000 lb. (55850 kg) compression and 24,000 lb. (10900 kg) shear. This tower was also designed to be, and was, erected by helicopter.
TRANSMISSION LINE TOWERS FOUNDA nONS
88
A series of five soil borings (Fig. 8) were taken as close to the transmission line centerline as :possible by using the rreandering canals, setting the drilling equipment off on the bank and taking a boring. This produced a profile that showeda layer of organic silt ranging from 25. ft. (7.7 m) at the south end to 75 ft. (33 m) at the Pamunkey River bank, with no strength, underlain by a dense very fine to fine sand.
\
\) ~
FIG.
6 - SCREW
ANCHOR
FIG.
7 - GRILLAGE
FIG.
9 - BASE
PLATE
The multi-helix screw anchor foundation designed for the Eltham Ha.rsh was a three anchor cluster, one vertical anchor and Th.D anchors battered (one transverse, one longitudinal), for each of the tOYler's four legs (Fig. 9). Each screw anchor consists of a 10 ft. (3 m) lead section 3.5 in. (76 mn) in diarreter, with three helices of varying diarreters, 10, 11.3, and 13.5 in. (254, 289, and 342 mn) on 36 in. (915 mm) spacings; a 10 ft. (3 m) extension section, 3.5 in. (76 rom) diarreter with two 13.5 in. (342 mn) helices; and the appropriate number of 10 ft. (3 m) length of 8 in, (203 rom) diarreter pipe extension to reach the required depth. Each anchor was to be installed to 10,000 lb ft (13560N m) of torque to penetrate the dense sand as far as :possible. There was no practical way of testing the anchors. After all anchors in a group were in place, they were cut off to grade, approxinately 24 in. (610 nm) above the marsh, and a 1.5 in. (38 nm) plate was set over the anchors and each anchor was welded to the plate. The tower base showis :positioned and welded to the plate. Tne helicopter guide angle was installed, (Fig. 10) and then the foundation was ready to accept the transmission tower.
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3LOW COUNT SLO ••••. S ~E:DU!RE:O ro ADVANCE THE DAMES \ ~aORE SOIL SAMPLE!=! .Q OISTANCE O~ Q".jE ~aOT USING A 140 THE [lA""~S G.O. OF Jlf.
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102
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TROWBRIDGE
90
TRANSMISSION
LINE TOWERS
FOUNDATIONS
The 230 kV line across the RoanokeRiver Swamp used ten 155 ft. (47 m) self-supporting double-circuit lattice towers, the same towers as on the Eltham Marsh line. This line was constructed on new right-of-way and access into the swamp for soil borings was very difficult. Wecut in from the land side approximately 300 ft. (100 m) to take one boring. Wecame in Broad Creek and cut both ways to get tv.Dborings, and from the river, we got one on the bank a!1d in about 150 ft. (45 m). A profile was then m:tdeof the three miles using the five soil borings (Fig. 11). This inforffi3.tionwas used to design the anchors. The profile showed 20 to 25 ft. (6 to 7.5 m) of peat, underlain by loose to firm fine sands which varied in depth at each boring from 12 to 30 ft. (3 to 9 m) and then into a stiff marine clayey silt. The anchors were designed to take up at 10,000 lb·ft (13560 N·m) of torque in the sand layer, at an approxim:tte depth of 50
ft.
(15
m) •
The anchor design was about the same as that used in the Eltham Marsh. Tne differences were: the extension section was a 6 in. (152 mn) diameter pipe instead of the 3.5 in. (76 mn) pipe and the pipe sections were joined together by 13.5 in. (342 rom)helix flanges bol ted together instead of threaded coupling. The rest or the threeanchor assembly was the same.
Fig. 10 Foundation For Lea of Lattice Tower
MULTI-HELIX
SCREW ANCHORS
91
Construction In the Dismal Swamp,the construction of the tower foundations was started approaching the swampfrom the west side. Union-Camp,a paper product company who owned the portion of the Dismal Swamp in NarlserrDnd,Virginia, at that tirre, had built sane pr:Lrnitiveroads for G~eir timbering operation. By using sane of these roads, the contractor could reach the transmission line right-of-way in the ce:1ter of the swamp. This center point and the ti_Dline-entry points into the swampwere the contractors only ground access for work. Because of the trafficability problems, the contractor used two tyt:€s of track equiprent along the right-of-way. The one used for the hea\y hauling of the foundation materials was a quad-track carrier called "Juggernunt" which is buil t li.."I(ea truck, but rW1Son about eiqht-foot tracks instead of wheels. The other type of equiprent was a small track vehicle called a "&::fT1bardier". Three different rrodels of this vehicle were used. Two m::x:.els,the "MuskegCarrier" and the "MuskegTractor", weighed about 7,000 lb. (3180 kg), travel on 28 in. (711 rom)wide tracks loaded with a r:B.Xi1m..rrn pay-load of 8,000 lb. (3630 kg), and have a zero penetration ground pressure of only 1.5 psi (10.3 ~~/m2). These tivo were used as personnel carriers and for light hauling. This type of vehicle was also used to make the soil borings. The third "Banbardier" vehicle was called the "Terrain Master", a four-powered track unit. The Terrain Master weighs 16,000 lb. (7260 kq) and can carry a maxi1m..rrn load of 15,000 lb. (6800 kg). wnen loaded, the ground pressure at zero penetration is 3.4 psi (23 kN/m2) for the power unit and 7 psi (48 kN/m2)for the loaded deck unit where the contractor mounted a hydraulic boom with a 10,000 lb ft (13560 N m), two speed rotating hydraulic digger unit. \'lith the above equiprent, the contractor approached the western edge of the Dismal Swampand started the foundation work for the forty guyed-Y aluminumtowers. His work procedure was to install the three base anchors--consisting of tivo 10 ft. (3 m), 3.5 :L~. (89 rom) hollow shaft pipe with increasing diameter helices, and then the 8 in. (203 rom)standard structural pipe--until the hydraulic digger unit registered a shaft torque of 9,000 lb·ft (12200 N.m)mini1m..rrn reading. The torque was measured by a dynamometerinstalled on the kelly bar beDoJeen the digger unit and the anchor shaft. The unit usually took b.D to four pieces of 8 in. (203 rrm)pipe. After all three units were in place, the 8 in. (203 rom)pipe was burned off to allow the top of the grillage to be 24 in. (610 nm) above the swamp. The foundation waS then ready to take the tower. The Terrain Master was positioned to install each of the multi-helix screw anchors at each guying point. The anchors varied in length from 28 ft. to 124 ft. (8.5 to 38 m) before reaching the required torque of 4500 lb·ft (6100 N.m). After each anchor was installed to the required torque reading, it was tested in tension to 25,000 lb. (11350 kg). The anchor was pulled to 5,000 lb. (2270 kg) and then loaded in
92
TRANSMISSION LINE TOWERS FOUNDA TrONS
increments of 5,000 lb. (2270 kg) with creep readings being made at each of the load increments thereby setting the anchor and checking alignment. Each tower fOllildationwas installed in the sameway. For the Eltham Marsh, the contractor set up a staging area on the north side of the PamunkeyRiver and movedmenand material across the river to where the marsh canal system rUJ1Sinto the river. He then followed the canal system into the marsh to each structure site. It was impossible to spot structure sites adjacent to the ca.'1als in all cases and getting from the edge of the canal to the construction site presented another major problem. Dredging and rrost conventional road building techniques for these conditions were eliminated by environmental restrictions. A search was llildertaken during the engineering stage fora practical means of getting fOllildation equipment to each tower site. Subsequently, a recorrmendationwas made to use a 12 x 21 ft. (3.7 x 8.2 m) sheet of laminate, called M~~T, a product of Air Logistics Corporation. These waffle-like panels appeared to have the qualities necessary to get men and equipment to the various tower sites without disrupting the delicate marsh ecol090'. The timber of the densely River was used from the road downthe center of the the three general access provided.
forested cypress swamp of the Roanoke right-of-way clearing to build a corduroy clearing. By using the corduroy road from points, access to each tower location was
For both the Eltham Marsh and the Roanoke River, the Bombardier "Terrain Master" with a hydraulic boommolliltedon the back was used to install the anchors. A two-stage hydraulic power head capable or providing 10,000 lb·ft (13560 N.m) of torque was ffiOlliltedto the boom to turn each anchor section into the grollild. Vertical anchors were installed first so that they could be used to aid in the installation. or the two battered piles. Each section of anchor was screwed into place leaving approximately 3 ft. (1 m) of pipe above grollild. The next section was then attached to the power head and placed on top or the protruding section and connected by bolting or threading. This process was repeated lliltil the specified torque was attained. The depth of anchors varied from 25 to 85 ft. (7.6 to 26 m) over the two projects. On the Eltham Marsh (Fig. 12), installation tiTre was approximately 30 minutes per section. Heavy root mass and underground obstructions in the RoanokeSwamp(Fig. 13) increased the installation tiTre considerably. Battered piles were started using predetermined horizontal and vertical distances for a 4: 12 triangle. Tne anchor waS set at an appropriate distance from its specified location and advanced vertically to a predetermined tip elevation. Tne anchor was then pulled horizontally to the specified ground line location to obtaill the proper batter. This rrethod worked satisfactori1.y in the Eltham Marsh, but the heavy root roass in the Roanoke SWampcreated many problems. Tnis rrethod waS selected by the contractor because of the swivel attachrrent of the power head to the boom. This non-rigid connection made it very difficult to start the 150 lb. (68 kg) lead
MULTI-HELIX
Fig. 12
Fig. 13
SCREW ANCHORS
Construction - ELTHAM ~~H
Construction - ROANOKE SW?~~
93
94
TRANSMISSION
LINE TOWERS
FOUNDATIONS
section on the proper batter without a rigid guide or template which the contractor elected not to use. The nature of the surficial soils on these projects allowed for the horizontal movementof the anchors to be accomplished satisfactorily, but soil with any appreciable shear strengr...hwould not allow the anchor to be movedhorizontally as in the arove rrethod. Consideration should be given to a rigid guide or template which would allow the anchor to be started at the proper batter with little horizontal movement. Unfortunately, the construction of the two projects was not problem free. The Roanoke Swampand its heavy root mass caused significant problems and construction delays. Subrnergedlogs and massive cypress roots made it very difficult to get proper anchor alignrrent. Muchof the debris had to be removed, this created significant loss of tirre. Anchor failure during construction occurred on both projects. On the Eltharn Marsh, two anchors failed near the cut off torque of 10,000 lbon (13560N,m). Sorreof the anchor sections were recovered, but actual cause of failure was undetermined. In the RoanokeSwamp, construction was plagued with several anchor failures. Tnese failures are thought to have been a result of the underground obstructions encountered during construction. The failures generally occurred at shallow depths and at relatively low torque. Wefeel that excessive horizontal movementof the anchors during installation put undue stress on the anchors and caused failure. Tne lead time for the anchor material was 8-10 weeks. Tnis created scheduling problems on the Earlej's Line when anchor failures occurred. When ordering material, anchor failure was not anticipated and no extra lead nor lead extension sections were purchased. Anchor penetration can be estimated, but actual length is not definite. Actual penetration of someanchors was 20 ft. (6.6 m) deeper than estimated.
In the Eltham Marsh, anchor penetration was within 3 ft. (1 m) of estimated depths and, except for the two anchors that failed, materials procurement was not a problem. Conclusion Surface and access conditions on all three jobs were the main factors for using multi-helix screw anchors. The peat and organic surface materials in the Dismal Swampmade the moving from site to site slow going. The delicate ecology of the Eltharn Marsh remains unblemished because we were able to use the lightweight equipment on the Ma1AT,and manpo.verto install a foundation capable of taking the tower loads. The ruggedness of the RoanokeSwampmade the going slow a.."1d took its toll in anchors. But, with such acces s limi ta tions as described aro\~ and increasing environrrental restraints, Virginia Power's engineers considered the high capacity multi-helix screw anchor foundation to be a viable alternative to conventional foundations.
MULTI-HELIX
95
SCREW ANCHORS
References "Encyclopedia
of
Anchoring."
Bulletin
1.
A. B. Chance Canpany, 4-7706, 1977, 29 pp.
2.
Joslyn Hardware Division, "Joslyn Power Installed Screw Anchor System HandJ:x:::ok."PLHD-PED-I-75;Chicago, Illinois, 1975, 121 pp.
3.
Rcdgers, T. E., Jr., "The Dismal Swampand the Successful 500 kV Line." CIGRE's (International Conference on Large High Tension Electric Systems), Study Carmittee 22 Meeting, Stuttgart, Gerrrany, JW1e 1973.
4.
Rodgers, T.E., Jr., and Elliott, O.F., Jr., "High Capacity of !v1ulti-Helix Screw Anchor for Self-Supporting To.ver FOW1dation." Southeastern Electric Exchange, Engineering & Ooerations Division, Transmission Section, May 9-11, 1979, Bal Harbour, Florida.
SPREADFOUNDATIONS IN UPLIFT: EXPERIMENTAL STUDY Fred H. Kulha~7l,
F.ASCE, Charles H. Trautmann 2, M.ASCE, and Costakis N. Nicolaides3 ABSTRACT
The uplift capacity of spread foundations can be influenced by the native soil density, backfill soil density, foundation depth; and foundation shape. Each of these factors was investigated for model spread foundations in dry sand by an experimental program of 90 uplift tests. Load- displacement data and observations of the failure mode were obtained, and the results indicate that backfill compaction increases the uplift capacity and stiffens the load-displacement response for all native soil densities, with greatest influence in dense soils. Foundation capacity increases substantially with depth, especially in dense soil. The test results agree well with published experimental studies in homogeneous deposits; however, there appear to be no comparative studies in which the densities of the native soil and backfill soil differ. These test results are relevant to the optimal design of foundations for electrical transmission line structures. INTRODUCTION Spread foundations are used extensively within the electric utility industry as the foundations for four-legged lattice towers. For example, the results of a recent survey by the Electric Power Research Institute showed that about half of all existing towers in the U.S., and about one-third of those planned for construction in the next decade, use this type of foundation (Kulhawy, et al., 1983). In spite of this extensive usage, many of the factors controlling the uplift behavior of these foundations are not understood adequately. It has been accepted for many years that the uplift capacity of spread foundations increases, in general, with increasing size and depth of foundation and increasing soil density. Field rest data also have been available that have shown the important effects of backfill compaction, for example as sho~~ in Figure 1. These data clearly illustrate that increased compaction of the backfill over the foundation increases the uplift capacity and stiffens the load-displacement response of the foundation. Experimental
and
analytical
studies
of uplift
and 3 Gradua te Iprofessor, 2Research Associate, School of Civil and Environmental Engineeering, Ithaca, NY, 14853-3501
96
capacity
largely
Research Assistant, Cornell University,
SPREAD
FOUNDATIONS
97
IN UPLIFT
125
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Displacement
0.8 m thick 0.4 OAm 0.6 0.2 thick layer, (m) Compacted c) Loose back fill 75 / ________ ~a
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Effect of Backfill (after Heikkal~
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Compaction on Uplift and Laine, 1964)
Behavior
have disregarded that spread foundations are constructed in excavations, and that the backfill can have a wide range of densities. Both the experimental and analytical studies have assumed, either explicitly or implicitly, that the native soil and the backfill are at the same density and state of stress and therefore are homogeneous. Unfortunately, this assumption does rot model the field case, in which the backfill can range from loosely dumped to very well compacted. The assumption of homogeneity therefore is only a special case of the general problem. In this paper, preliminary results are presented of an extensive laboratory study of the behavior of model spread foundations in uplift. In these tests, the foundation size and depth have been varied over typical ranges employed in practice, and the field construction process has been simulated from excavation through backfilling, using a range of densities. The general behavior observed in these tests is presented herein. TEST FACILITIES An overview of the test apparatus is shown in Figure 2. All of the tests were conducted in a chamber fabricated from a standard 210 liter steel drum. Each test was prepared individually and, in every case, the observed uplift failure surface was located well aMay from the walls of the chamber. The lID de 1 foundations rreasured 100 by 100 nm and 100 by 200 11m and were fabricated from 6.4 mm thick steel plate. A 6.4 11m rod was threaded into the center of each plate to transfer the uplift force from the loading system to the plate. The weight of the model foundation was subtracted from the gross measured force during data
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SPREAD
FOUNDATIONS
IN UPLIFT
99
reduction. The soil used in the tests was a mixture of a filter sand and a silty fine sand available near Ithaca, NY; both materials are glacially derived. The filter sand is sub-angular outwash material containing limestone, quartz, and other rock fragments, and the silty sand is a lacustrine material containing mostly quartz. A grain-size curve for this composite soil is shown in Figure 3, and the results of direct shear tests are shown in Figures 4 and 5. 'l"..lenty three direct shear tests were conducted over normal stress levels of 2.5 to 25 k.."tjm2, which correspond to the range of normal stresses in the actual tests. Additional information on the soil properties and test methods is given by Nicolaides, Kulhawy, and Trautmann (1987). The uplift loads were applied to the rod e.xtending from the center of the foundation by a standard roller chain. This was gear-driven by an electric rrotor at a loading rate of approximately 2 rnrnjmin. These loads were monitored by a load cell having a precision of about 5 N. Displacements were monitored by a DCDT having a TIm. All readings were made using a precision of about 0.2 Hewlett-Packard HP-3455A multimeter under the control of a HP-9825A desktop computer. LXPERIMENTAL PROGRAM A total of 90 tests were performed, in which the variables were the ratio of foundation depth to width (1, 2, and 3), ratio of foundation length to width (1 and 2), native soil density, and backfill density. In designing the test program, emphasis was placed on modeling the actual field construction procedure as closely as possible to ch.lplicate the stress history that occurs in practice as a result of excavation, construction, backfilling, and loading. For each test, the native soil was placed by one of four different procedures. Then a hole was excavated, being particularly careful to avoid disturbing the native soil. The model foundation was placed in the excavation and then backfilled by one of three different procedures. The procedures used and measured soil densities are presented in Table 1. No correlation was made to relative density because the placement procedures were different. However," loose" material was placed by carefully releasing the soil from a small scoop, using a drop height of less than 100 run. Hedium-dense material was created by placing loose soil as described above and then compacting it with a 60-Hz electric vibrating plate. Dense soil was created by one of two procedures: the first employed strong vibration, while the. other employed a falling weight on a plate resting on the soil surface. In the loose native soil, it often was difficult to maintain the excavation walls, so they were "stabilized" by spraying a fine mist of water to establish a capillary stress or by using a square or rectangular sheet metal casing. The water mist did rot penetrate into the native soil more than 1 to 2 rrm. For rectangular foundations at the greatest depth, partial collapse of the excavation walls required casing for support. In these cases, backfilling was done and the'
TRANSMISSION LINE TOWERS FOUNDA nONS
100
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was placed in layers.
The testing program consisted of a partial factorial or parametric experimental design, generally with one replicate for each combination of parameters. The test variables ~re ordered randomly to eliminate any possible systematic effects of long-term changes in
SPREAD
Table 1. Soil
FOUNDATIONS
Soil Densities
P1uviation 2.1 0.42 Gentle 65 62 vibration S.D.b Placement 4.1 27 2.9 b - standard ofdeviation 20.35 19.79-20.99 1.7 0.35 80 Na 20.31 19.45 17.94 18.26-20.41 16.86-18.63 19.17-21.26 Loose 2.6 2.0 0.50 0.36 92 tamping 17.16 16.18-18.85 0.71 19.05 19.93 17.94-20.47 18.85-21.05 0.56 0.51 32 Mean c - coefficient variation Technique Range Heavy Tamping Strong vibration Light tamping COVc
IN UPLIFf
101
Measured in Tests (%)
Density (kNfm3)
Condition
apparatus during the course of the testing. Some combinations of parameters were not included, particularly those involving rectangular shaped foundations. For these cases, a general trend was determined on the basis of a limited number of tests on rectangular models. LXPERIMENTAL RESULTS The principal data from the tests consist or load-displacement curves and observations of the failure rrodes. These data show a number of trends that have significant implications for design practice. A summary of the key results is presented below. Load-Displacement
Response
The general pattern of the load-displacement curves is shown in Figure 6. As indicated, the response of the foundation becomes increasingly dilatant as the soil density increases. Concurrently, the foundation capacity increases, with the amount of increase being a function of the foundation depth and shape, as well as the soil density. In each case, however, the capacity at large displacements, soil when mrmalized by the factor iDBL, in which i = backfill density, D = foundation depth, B = foundation width, and L independent of the initial foundation length, appears m be relatively soil density. Furthermore, as the peak foundation capacity increases, there is a tendency for increased stiffness in the load-displacement response. This finding is important for practice, since the limiting factor for spread foundations in uplift commonly is displacement, rather than ultimate capacity. Failure
Mode
Three failure modes ~re observed, including shear along vertical surfaces extending upward from the edges of the foundation, ~dge or combined wedge and side shear failure, and punching failure. Most of
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Load-Displacement
Curves
the tests exhibited failure by shear along vertical surfaces, as illus trated in Figure 7. Wedge or combined shear failure occurred, in general, for foundations with DjB less than two in medium to dense native soil, where the backfill was at least 85 percent as dense as the native soil. This failure mode is illustrated in Figure 8. Punching failure occurred only at DjB equal to three where the backfill was less dense than the native soil. Punching failure produced essentially no disturbance at the soil surface as the soil near the foundation flowed down around the edges of the foundation model. In practice, spread foundations for transmission structures are rarely buried deeper than DjB = 3, and this depth ratio was the maximum used in the tests. Based on observations in previous studies, punching failure would be the tendency for foundations as DjB increased beyond about three (e.g., Esquivel-Diaz, 1967). Effect
of Backfill
Density
the Increased backfill density was found to increase portion of capacity and the stiffness in the initial This figure displacement curve, as shown in Figure 9. for the three uplift load as a function of displacement v.rith DjB equal densi ties, using the square model foundation The loose and dense native soil cases are shown in (a) respectively.
foundation the loadshows the backfill to three. and (b),
For the loose native soil, densifying the backfill increased the capaci ty by about 40 percent, while the displacement required to reach 50 percent of the capacity (corresponding to a typical design factor of safety of two) decreased by 75 percent. For the dense native soil, the effect of densifying the backfill was to increase the capacity by about 110 percent, while the displacement required to reach 50 percent of capacity decreased by 35 percent. These effects rended to decrease at shallower depths. Effect
of Native The
native
Soil soil
Density density
also
had a marked effect
on foundation
SPREAD
FOUNDATIONS
103
IN UPLIFT
o Backfill Side 1\t
Shear
Native
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Figure
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b)
Plan View of Failure Surface
Surface
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Mode Observations
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Mode Observations
capacity, with the effects being more pronounced at greater depths and where the backfill was ~ll-compacted. This behavior is illustrated in Figure 10, which shows the load-displacement response for square model foundations with loose and dense backfill. The capacity increased about 190 percent as the native soil density increased from loose to dense with loose backfill. For densely compacted backfill, the increase was about 365 percent. These results indicate that there is significant interaction between the native soil and backfill, and that both need to be addressed in design. The results also indicate that the effect of compaction is rruch greater in dense native soil. For sites with dense soil, backfill compaction can lead to very large increases in capacity which may outweigh the costs of ~eper or larger foundations.
TRANSMISSION LINE TOWERS FOUNDA nONS
104
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Influence of Backfill
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Density on Load-Displacement
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Density
Displacement,
20
mm
mm
Influence of Native Soil Density on Load-Displacement Response
SPREAD FOUNDATIONS IN UPLIFT
Effects
of Foundation
105
Depth
Figure 11 shows the effect of foundation depth for a square foundation in both loose and dense native soil wi.th varying backfill. In this figure, the net foundation capacity has been rormalized by the factor -yDBL (after subtracting the foundation weight). As shown, depth has a maj or effect on capacity for dense native soils, with increases up to 500 percent. For loose native soils, the effect is smaller, with increases up to 75 percent. Effects
of Foundation
Shape
Spread foundations for transmission line structures commonly are square, although rectangular foundations are used occasionally; in these instances, LIB ratios generally are less than two. Several tests were performed to evaluate the effect of shape, and it was found that the square foundations tended to have a higher dimensionless capacity factor than the rectangular counterparts at the same D/B ratio. The few exceptions to this general observation appear to be from random experimental errors. The effect is greatest for dense native soil deposits. This finding does mt imply, however, that square foundations have higher capacities than rectangular foundations of the same area at the same depth because, in this case, the rectangular foundation has a smaller width B and a correspondingly greater D/B ratio. When correction for this is made, the data indicate that there is little, if any, difference between the capacities of square and rectangular foundations of equal area at equal depths. COMPARISON WITHPUBLISHEDEXPERIMENTAL STUDIES There are few data in the literature for uplift tests on rrodel foundations in which the densities of the native soil and backfill soil differ; most published studies have been performed by placing model foundations or anchors on a soil surface and then placing layers of soil above. For these studies, the "native" soil and "backfill", as defined in this study, would be identical. These published studies can be compared with the present test results in v.hich the native soil and backfill v;ere placed at the same density. The results will still be influenced to some degree by the excavation procedure and the accompanying changes in stress; however, the effects should be relatively small for carefully prepared soil deposits. Figure 12 shows several published test results plotted with those of the present tests. The results agree reasonably ·~ll. The results for dense soil are in close agreement with the results of Esquivel-Diaz (1967), which were conducted in dense sand. The results for medium dense sand fall slightly below those of Baker and Kondner (1966), Clemence and Veesaert (1977), and Balla (1961) conducted in medium dense sand. They fall slightly above those of Das and Seeley (1975) reported for loose sand with a friction angle of 34 degrees. It
is
difficult
to
evaluate
the
noted
differences
precisely,
0Q.'" •....
uD/B
ClJ
CD - 0 0I
•....::>
S
ure 11.
20
-'
347 btackfill •backfillI•D/BSquare Loose backfill •"-Legend: •Legend: Medium dense backfill Dense IVariation 234IIDepth, a00u00>- ~u'"0106 native, of UpliftLINE Capacity with Depth TOWERS Factor FOUNDATIONS c: E Dense native,TRANSMISSION Square "- u •3"'"cc:>-:0 5(a)10(b)01615250I Loose
....... "Vi
-'
ClJ l.L. Q..l:> "in .l:> "~CD
t 25[
"-~ .~:~
I~. D,,,, b,,'WI
A/·
.~;
because test results are influenced by a large nW11berof factors, such as soil type, soil density, soil strength characteristics, and scale of the tests. In particular, the frictional strength data reported for other studies are generally not accompanied by descriptions of the
SPREAD
u
...en C1J
--
16 u~. 24 12 48 a0c0Eca0"3I..:0 20 >. C1J l.L ;.... <.) eC nD en
FOUNDA TrONS IN UPLIIT
107
28
"-
..0
..J
0
SQUARE
MODELS:
Yb
+-+x Medium dense Dense - tamped
This
X*-* 0-0 0- -0
Dense-vibrated Loose } Baker and Kondner
.--.
ESQuivel-
[).--[).
Dos and Seeley Balla (1961) Clemence and Veesaert (1977)
A--A
0--0
Diaz
Yn
=
study
(1966)
(1967) (1975)
2 Dimensionless
Figure
12.
3 Depth,
4
0/8
Comparison with Other Experimental
Studies
type of soil strength test and the mrmal stress levels used in the tests. The latter can have a significant effect on the measured friction angle for granular soils, as shown clearly by the stress dependency indicated in Figures 4 and 5. Given these uncertainties, the agreement between the experimental results appears to be good. DISCUSSIONANDDESIGNIMPLICATIONS The test results have a number of implications for the design of spread foundations for electric transmission line structures. First, the effect of compacting the backfill is to increase the uplift capaci ty and the stiffness of the uplift response to loading. This effect is greater at sites where the native soil is relatively dense. For projects in which excavation costs and/or foundation fabrication costs are greater than compaction costs, it is therefore more economical to compact the backfill well than to require larger or deeper foundations. Second, the uplift capacity is a function of both the backfill by tests in loose and native soil. This effect is shown clearly native soil with differing backfill densities. Therefore, compacting
TRANSMISSION
108
granular the soil
native soil is loose.
LINE TOWERS
always will
provide
FOUNDATIONS
beneficial
effects,
even when
Third, the uplift capacity of spread foundations increases dramatically with depth as D/B increases from one to three. This effect is particularly evident when the native soil is dense, because the soil dilates during shear and is better able to mobilize the strength of the soil mass above the foundation. SUMMARY This paper has described an experimental study of the effects of native soil density, backfill density, foundation depth, and foundation shape on the uplift capacity of model spread foundations in dry granular soil. The results are summarized in Table 2, which indicates the relative effects of parameter increase on capacity. Because the factors are interdependent, it is not possible to specify the effect of one parameter without first indicating the values of the others. In general, the results show that increased backfill compaction and increased foundation depth lead to significant increases in foundation capacity, while foundation shape has relatively little, if any, influence. Table 2. Increase in Parameter
Qualitative
Trends
Effect on Capacity
in Uplift Conditions in Capacity
Capacity for Which Change is Most Pronounced
Backfill Density
Increase
Deep (D/B = 3), Dense Native Soil, Square
Native Soil Density
Moderate increase
Deep (D/B = 3), Dense Backfill, Square
Depth
Substantial increase
Dense Native Soil and Backfill, Square
(D/B)
Length (L/B)
Little, if any, increase
ACKNOwLEDGMENTS This study was sponsored by the Electric Power Research Institute under Project RP1493-4, for which Vito J. Longo was the EPRI Project Manager. Appreciation is extended to Paul Jones and Glenn Darling, who fabricated much of the experimental apparatus, to Lorraine Crouse, who typed the text, and to Ali Avcisoy, who drafted the figures.
SPREAD
FOUNDATIONS
IN UPLIFT
109
REFERENCES 1.
Baker, W. H. and Kondner, R. L., "Pullout Load Capacity of a Circular Earth Anchor Buried in Sand", Record 108, Highway Research Board, Washington, 1966, p. 1-10.
2.
"The Resistance Balla, A. , to Breaking Out of Mushroom Foundations for Pylons" , Proceedings, 5th International Conference on Soil Mechanics and Foundations Engineering, Vol. 1, Paris, 1961, pp. 569-576.
3.
Clemence, S. P. andVeesaert, C. J., "Dynamic Pullout Resistance of Anchors in Sand", Proceedings, International Symposium on Soil Structure Interaction, Vol. 2, University of Roorkee, 1977, 31 p.
4.
Das, B. M. and Seeley, G. R., "Breakout Resistance of Shallow Horizontal Anchors", Journal of the Geotechnical Engineering Division, ASCE, Vol. 101, No. GT9, Sept. 1975, pp. 999-1003.
5.
"Pullout Esquivel-Diaz, R. F., Anchors in Sand", Soil Mechanics Durham, NC, 1967, 57 p.
6.
HeikkaUI, K. and L:iine, J., "Uplift Resistance of Anchor Plates", Proceedings, 20th Session of the International Conference on Large Electric Systems at High Tension (CIGRE), Vol. 2, Report 217, Paris, June 1964, 14 p.
7.
Kulhawy, McGuire, Structure EL- 2870, 1983, 412
8.
Nicolaides, C. N., Kulhawy, F. H., and Trautmann, C. H., "Experimental Investigation of the Uplift Behavior of Spread Foundations in Cohesionless Soil", Report El-xxxx, Electric Power Research Institute, Palo Alto, CA (in press).
Resistance of Deeply Buried Series No. 8, D.1ke University,
F. H., Trautmann, C. H., Beech, J. F., O'Rourke, T. D., Line W., Wood, W. A., and Capano, C., "Transmission Foundations for Uplift-Compression Loading", Report Electric Power Research Institute, Palo Alto, CA, Feb. p.
Uplift of Shallow Underreams
in Jointed Clay
Azaroghly Yazdanbod,l Shamim A. Sheikh,2 and Michael W. O'Neill, M. ASCE3
ABSTRAcr Four full-scale belled footings with nominal depth-to-bell-diameter ratios in the range of 1.0 to 1.67 were tested in uplift in a deposit of naturally occurring overconsolidated, desiccated clay. The footings were instrumented to permit separation of soil suction from frontal soil resistance and were subjected to rapid monotonic, sustained monotonic and cyclic loading. The effects of soil suction, which developed as a result of negative pore water pressures in the soil, were found to be significant at large deflections but not to be maintainable at a large magnitude for long periods of time. Sustained loading did not significantly affect the footing capacity; however, one-way cyclic loading reduced the uplift capacity significantly. The results of the tests are modelled by a simple mathematical equation.
INTRODUCTION Transmission line towers and other tower structures are often subjected to lateral shears and overturning moments sufficient to produce significant uplift loads on their foundations. A number of foundation systems can be used to resist such loads, including footings with sufficient dead weight to completely balance the applied load, deep foundations to resist the load mainly through side shear on their shafts, helical anchors, shallow dug footings and shallow belled footings. The latter type of foundation, which is the subject of this paper, is applicable when cohesive soils exist near the surface with sufficient mass strength to permit the formation of a bell without the use of drilling fluids. In some cases they may be preferable to shallow, straight-sided piles, which could lose a portion of their side shearing resistance during repeated storm loadings. A shallow belled ( or "underreamed") footing can resist uplift load through as many as three distinct mechanisms ( 3 ), as described graphically in Fig. 1 and symbolically in Eq.l: Tu=W
(1)
+Q+S,
1 Graduate Student; 2 Associate Professor; and 3 Professor; Department of Civil Engineering; University of Houston - University Park; Houston, Texas 77004
110
III
UPLIIT IN JOINTED CLAY
-: -Shaft I I
w
~
j
I I I
I I
Arbitrary limit of soil included in W
I I
Bell
I I
Reamer seat
j S
Fig. 1. Schematic of a Belled Footing Under Uplift Loading su(ksf)
o
o
1
---•
3
2
• su= 1.08 + 0.160(D(ft)
-
- 2) ksf
•.......
•
'""
.!::.
Q
---
Bottom of Footing B
---
Bottom of Footings A,C,D
Q)
o
10
su = 1.88 + 0.083 (O(ft)
•
-
n
k sf
•
16 Fig. 2. Undrained Shear Strength Vs. Depth; UU Triaxial Data (1 ft = 0.305 m; 1 ksf = 47.9 kPa)
112
TRANSMISSION LINE TOWERS FOUNDATIONS
where T u = total ultimate uplift resistance, W = appropriate weight of the footing and some zone of overlying soil, Q = frontal resistance of the soil acting downward on the top of the bell and S = suction (tension) developed between the soil and the bottom of the footing. The force Q is the vertical resultant of the shearing stresses that develop on the failure surface in the soil above the base of the footing. [Q may act partially in shaft friction if adequate bonding exists between the soil and the shaft. This will be shown not to be the case for the shallow footings considered in this study.] To develop a rational procedure for the design of belled footings under uplift loading, it is important to evaluate these components separately for conditions of geometry and loading that are typical of in-service foundations. This paper describes a series of instrumented, full-scale, footing tests conducted in moderately jointed, saturated, overconsolidated clay, in which two depth-to-bell diameter (DIB) ratios were studied (approximately 1.0 and 1.67) and in which loads were applied as rapid monotonic, sustained monotonic and one-way cyclic axial forces at the tops of the footings. The tests were conducted to provide full-scale data to assist Houston Lighting and Power Company's evaluation of design procedures based on smaller-scale uplift tests ( 2, 8 ); however, this paper does not address design procedures.
GEOTECHNICAL
CONDITIONS
The site of the uplift tests was the University of Houston Foundation Test Facility, located in Houston, Texas, about 3 mi (5 km) southeast of the downtown district. The general geological and geotechnical conditions, as well as the behavior of one shallow footing tested in compression at this facility, are well documented ( 4, 6). The near-surface soil belongs to the Beaumont Clay formation, a Pleistocene-aged plastic clay that was preconsolidated by a process of desiccation that left the soil with a network of closed, discontinuous joints. Profiles of undrained shear strength at the location of the footing tests, as measured with UU triaxial compression tests and quasi-static CPT tests conducted with a one-piece elecronic cone penetrometer, are shown in Figs. 2 and 3, respectively. UU triaxial test samples were taken using thin-walled tube samplers from 3 borings adjacent to the test footings, and 6 CPT soundings were made in a matrix pattern throughout the location of the footing tests. The highest water table at the site is located at a depth of 7 ft (2.13 m); the overconsoldation ratio of the soil is about 8 at a depth of 10 ft (3.05 m) (the shallowest depth at which OCR could be reliably measured); and the average plasticity index and total unit weight of the soil above a depth of 10 ft (3.05 m) are 30 and 126 pef (19.8 kN/m3), respectively. Zero shear strength is indicated above a depth of 2 ft (0.61 m) on the profiles of undrained shear strength. This was the depth to which surface joints and brittle, highly desiccated soil were observed to penetrate and which were presumed to render the soil ineffective in providing frontal uplift resistance against the bell (Q) or shear resistance between the shaft and the soil mass. The qc values from the quasi-static CPT were converted to undrained shear strength by first subtracting total vertical stress and dividing the result by 19, a correlation factor that has been developed for the test facility (4). The spikes in the CPT profiles were the result of calcareous nodules and occasional sand seams that are not effective in providing uplift resistance. Tnerefore, the interpreted shear strength profile was drawn to eliminate the spikes.
UPLIFf IN JOINTED CLAY Su
o
2
4
(ksf) 6
o
5
-S
10
o
15
113
8
10
12
Su =
1.20 + 0.140(D(ft) - 2) ksf
Su =
1.90 + 0.245(D(ft) - 7) ksf
£. •... 0. Q)
Su=
20 25
Fig. 3. Undrained Shear Strength Vs. Depth; CPT Data (1 ft = 0.305 m; 1 ksf = 47.9 kPa)
OrO
E 0'-".t= OJ
-a. -0a. •....•
•....• .•.... .t= OJ
Footing (D/B
=
B
0.99)
~ NI ...... I ••• N ~ C') 0 OJ N r-N r--
aJ
~l
o
r--
Footings A,C,D (1.56~D/8~1.69)
N
, I,I II!00.31 I 0.38 I1.00 1.25 0.31 1.25 I 1.00 0.38 3.00 0.92 Distance (tt) Distance (m) ,
1.16 3.79 I
I
r
3.00 L...J
3.79
0.92 L...J
1.16
Fig. 4. Geometry of Test Footings
114
TRANSMISSION
LINE TOWERS
FOUNDATIONS
TEST FOOTINGS AND TEST PROCEDURES Profiles of the four test footings are shown in Fig. 4. Each of the footings was machine-excavated in the dry in approximately 60 min and was concreted within one hour thereafter with 3- to 6-in. (7S-1S0-mm) slump concrete having an unconfined crushing strength of approximately 6000 psi (59 MPa) at the time of the footing tests (approximately 120 days after construction). A full-depth reinforcing cage, consisting of 8 No. 10 deformed bars longitudinally and No.4 deformed bar hoop reinforcement at an 8-in. (200 mm) pitch, was installed in each shaft. A separate lifting apparatus, consisting of high-strength steel bars bolted to an anchor plate cast inside the cage immediately above the bell, extended out of the footing to a jacking point several feet above the top of the footing. Load was applied by jacking upward against a yoke that was attached to these high-strength bars. The jack rested on a pair of reaction beams that were in turn supported at their ends by surface mats located about 12 ft (3.7 m) away from the center of the test footing. Pressure was supplied to the jack by an electronic pump, and load was measured by an electronic load cell placed between the jack and yoke. Deflections of the top of the shaft of the footing were measured by four dial gages suspended from reference beams aligned perpendicular to the reaction beams. The reference beams were supported on posts driven into the ground about 12 ft (3.7 m) away from the center of the footing. The dial gages were placed on the perimeter of the shaft at 90-degree angular spacings to permit rotational effects to be observed and cancelled, if necessary. Radial lines of survey monuments were also established beginning on the east and west sides of the footing and extending outwards along the ground surface between and parallel to the reference beams at 1.0- to 1.5-ft (0.305- to 0.46-m) intervals to points immediately adjacent to the support posts for the reference beams. Optical surveys were performed throughout the tests using a stable backsight on a distant deep, massive drilled shaft to (a) confirm that the reference beams were not moving (which was found to be the case for all tests for all practical purposes) and to (b) obtain surface profiles of soil deformation. In order to minimize the effects of the thermal environment and rain on the readings, the site of each test was covered with a large tarpaulin. ~ Both total and pore water suction was measured directly beneath each footing. The total suction was measured by suspending an air pressure sensor inside a small, plastic-lined, 8-in.- (200-mm-) deep cavity that was hand carved at the base of each reamer seat. Pore water pressure was measured by embedding a saturated pore water pressure transducer in the saturated clay 12 in. (300 mm) below the base of the reamer seat. Both of these transducers, which measured pressures positively or negatively relative to atmospheric pressure, operated on the vibrating wire principle and were therefore relatively unaffected by moisture intrusion. Electrical resistance strain gages were placed on the reinforcing cages at the top of each bell and the top of each shaft to measure the shear load transfer in the shafts. Unfortunately, moisture apparently penetrated the waterproofmg during the time between casting and testing, rendering the strain gages ineffective. It was therefore necessary to estimate the shear load transfer in the shafts by indirect means. The testing sequence was established to provide a loading protocol that was representative of loadings that are applied to transmission line tower foundations:
UPLIFr IN JOINTED CLAY
115
monotonic loading (rapid, as occurs if unbalanced line tension develops during construction, and slow, as occurs when permanent unbalanced loads exist on towers) and cyclic loading, such as may be developed by high wind gusts or seismic events. The tests were conducted as follows. Footings A and B (nominal DIB of 1.67 and 1.0, respectively) were tested in rapid monotonic uplift to failure (defined as continuous upward movement under constant load) in a time period of approximately 50 min., with uniform load increments being applied every 5 min., after which they were unloaded and reloaded again to failure in approximately the same period of time to observe post-failure behavior. Footing B was also subjected to a third cycle of loading. Difficulty was experienced with the suction pressure recording device during the test on Footing B, so no usable suction data were available for that footing. Footing C was tested under monotonically increasing sustained loads of 18, 36, 54, 72 and 90 percent of the ultimate uplift capacity of its geometric twin, Footing A. Loads were maintained for 48 hr without unloading for each of the former four loads. When the fifth load was applied, steady upward movement was observed, so that the load was held for only two hI. Following the sustained load test, Footing C was also subjected to a rapid monotonic test in a manner similar to the tests on Footings A and B. Footing D was subjected to one-way cyclic loading at load amplitudes of 32 percent (125 cycles) and 55 percent (100 cycles) of the ultimate capacity of Footing A, which was geometrically identical to Footing D. The cycling period was 3 to 15 min. Following the tests at the second load amplitude, the load amplitude was decreased again to 32 percent of the capacity of Footing A, and 15 cycles were applied to investigate the effect of cyclic movement at low amplitude loading following cycling at a higher load amplitude. Finally, the load amplitude was increased to 73 percent of the capacity of Footing A, at which time the footing failed after the application of 6 cycles of load. Additional information can be found in Ref. 7.
TEST RESULTS Footin g A The "baseline" test was the test conducted on Footing A; therefore, its behavior will be described first. The load (T)-total suction-uplift deflection data are shown on Fig. 5. Failure of this footing first occurred at a deflection of about 2.0 in. (50 mm) (about 3 percent of the bell diameter) and a load of 220 K(980 kN). Total suction could not be measured accurately at that load, but extrapolation of the first-cycle suction-load curve to 220 K (980 kN) and observation of the second-cycle suction-load curve suggests that its value was about 3 psi (20.7 kPa). Loading in the second cycle was carried out to a total displacement of 7 in. (175 mm) with no decrease in total load but with an increase in ma.ximum total suction to about 10 psi (68.9 kN) at maximum deformation. Pore water suction is not shown on Fig. 5 for clarity, but it was found to track the total suction almost identically, indicating that loading did not produce soil framework (effective) stress changes directly below the base of the footing. The suction was a time-dependent variable, decreasing by a factor of about 2 from the maximum value measured directly after applying a load to the end of the 5-min. hold period between load applications. Time-dependent suction decrease in these tests is thought to be associated with the development of minute pathways in the jointed clay between the atmosphere and the base of the footing. It is unlikely that it was the result of air coming out of solution in the pore water and diffusing through the plastic liner in the total suction cavity because the liner used had a high air entry point.
116
(/)
~
•...
TRANSMISSION LINE TOWERS FOUNDATIONS
-....-
•-
-
Co
c: c: :J.-
----- --
increment applied 00- o:Jo(/)U)c:U(t\(t\ U)U)465320=.I 575 min. 10 .r. r-_ ............. a tter........ load&:__ ~ .... T (K) ....•.~ -_ ....•~•..•....Cycle applied ~ increment f-Q: u +30 ~;;: '~er load .~ Cycle 2 250 (J)
(J)
(J) (J)
1
-r
Fig. 5. Load Vs. Deflection and Total Suction, Footing A (1 K = 4.45 kN; 1 in. = 25.4 mm; 1 psi = 6.89 kPa)
0c:
.r.
~ 0(/) c: u (J)
27
(J)
1
--
~M,"'"'"
(t\
Measured
T
(T-S'
6r
50
100
150 200
250
T (K)
Fig. 6. Actual and Corrected Load Vs. Deflection, Footing A (1 K = 4.45 kN; 1 in. = 25.4 mm)
UPLIFf IN JOINTED CLAY
117
It is also apparent that total and pore water base suction were functions of displacement, with suction pressures approaching one atmosphere being devloped only after large upward displacements of the footing. It may also be inferred that the frontal resistance of the soil above the bell (Q, Fig. 1) decreased during the large deflections applied in Cycle 2 from the fact that suction was much larger in Cycle 2 than in Cycle I while the total capacity remained constant. To illustrate that effect, a graph of total load minus suction load (total suction pressure times base area) versus deflection is compared with the total load versus deflection curve in Fig. 6. The limiting total load minus suction was approximately 179 K (797 kN). Surface deformation patterns are shown on Fig. 7. Ground surface deformations approached zero at distances of greater than 100 in. (2.5 m) from the center of the footing, and a slightly smaller slope. occurred on the ground surface directly above the bell than at distances beyond the horizontal limits of the bell. The interpreted exit point of the failure surface is shown in Fig. 7. It was also observed that the shaft movement was discontinuous radially with the ground movement, suggesting that the soil is not bonded to the shaft. A distinct surface fracture pattern also developed during loading. The pattern of fractures on the surface that were mapped at the conclusion of the second cycle of load is shown in Fig. 8. The tangential crack around the collar of the shaft appeared fIrst at a load of 120K (530 kN), followed by the tangential crack on the west side of the footing at a distance of 55 to 70 in. (1.5 to 1.8 m) from the center of the footing at a load of 140 K (620 k1~). The latter crack was within the visible uplift zone of soil and was apparently caused by tensile strains in the soil, as its geometry is not consistent with the natural surface joint pattern. The radial cracks appeared at loads of 140 to 200 K (620 to 890 kN). The soil was also visibly pulled away from the lateral surface of the shaft to a depth of at least 4 ft (1.2 m). This observation, in conjunction with the observation regarding the discontinuity of shaft and ground surface deformations, indicates that essentially no load was transferred in side shear at the time of failure. The observations described in Figs. 7 and 8 suggest the failure mechanism shown in Fig. 9. That is, the uplifted soil appeared to be confined to a solid body approximated by a truncated cone, with an apex angle of about 27 degrees with the vertical. Failure was clearly influenced by the presence of the ground surface, and the mechanism is obviously "shallow" rather than "deep." However, since the sum of the weight of the soil and concrete inside this solid body and the suction acting at its base do not approach the value of the load applied at failure, significant shearing resistance was apparently developed along the surface of the body at failure. The force Q in Fig. 1 is the resultant of this resistance, which, as may be inferred from Fig. 6, reduced somewhat with increasing deflection after its peak value was reached.
Footing B The load-deformation curves for the three loading cycles for Footing B (DIB = 1.00) are shown in Fig. 10. The maximum load in Cycle 1 was 109 K (485 kN), at which time the deformation began to increase very signifIcantly. The load was maintained at this level to check for possible structural or jacking system failures for a period of 20 min., instead of the usual 5 min., after which a deformation of nearly 4 in. (100 mm) was reached. While the suction recording device did not function during this test, the rapid deformation during the maintenance of the 109 K (485 kN) load was interpreted to
118
TRANSMISSION
LINE TOWERS
FOUNDATIONS
Surface Deflection (in.) • T = 100K.
Cycle
1
o T = 180K.
Cycle
1
T = 220K. Cycle 1 + T = O. End of Cycle
I:>
Interpreted Failure Plane Exit
o T = 220K.
Cycle
2
~
5
5
10
Distance west from shaft face (ft)
10
Distance east from shaft face (t1)
Fig. 7. Deflections on Shaft and Soil Surface, Footing A (1 in. = 25.4 rom; 1 ft = 0.305 rn)
I
I
I
/
/'
./
.--
Crack
width exaggerated
I
/
\
\
/
~
Bell outline
"--_.--/ ./ H /
1 ft (0.305m)
Fig. 8. Surface Crack Pattern, Footing A
1
UPLIFT IN JOINTED
119
CLAY
A
limits of truncated
cone
Probable true failure surface I. 5 It .1
1.53m
Fig. 9. Interpreted Failure Surfaces; Footings A and B
:J
.c c:: c: ()
0 a(/)
362 9 5 7
;1)
8
(Ij
50
100
150
T (K)
Fig. 10. Load Vs. Deflection, Footing B (1 in. = 25.4 mm; 1 K = 4.45 !eN)
• T = 80K. Cycle 1 o T = 100K, Cycle 1 o.T=109K,Cycle 1 (20 min.) + T = O. End of Cycle 1
o T = 130K. Cycle 2
I
10
Fig. 11. Shaft and Soil Surface Deflections, Footing B (Iio. = 25.4 mm; 1 ft = 0.305 m)
TRANSMISSION
120
LINE TOWERS
FOUNDATIONS
3
0 ()
Q) .•...
c:
-
0~:; -~c:
~
1
.•... Q) CJ)
120K T= •ooT=190K TT == 160K 80K
+ T = 40K
Do
J
21
o 0.1
~
e
1.0
;
10
••
r •..•.•.•••
Time (min)
100 1,000
10,000
o •.....
c: .-
-
.-o •...... 0. (/)
1
() Q) ::J •... ::J
CJ)
cu -
~
o Q)•...
2
1-0..
Zero
suction
measured
for T = 40K
3 Fig. 12. Deflection-Total Suction-Time Relationships, Footing C (1 K = 4.45 kN; 1 psi = 6.89 kPa)
0c:
~Q)0
-~0 432501 ()
.•... Q)
-eT
•.....
c:
•...... CJ) .•...
---0
T -
+T:
50
100
150
200
S
Cycle 1, Footing A (transla ted)
250
T (K) Fig. 13. Load Vs. Deflection, Footing C (Reload) and Footing A (1 in. = 25.4 mm; 1 K = 4.45 kN)
UPUFf IN JOINTED CLAY
121
be due to the release of suction. The footing was then unloaded and reloaded more rapidly than in Cycle 1 (in 15 min.), and an increase in capacity was observed. The deformation associated with the second-cycle reload was about 5 in. (125 mm), and it is hypothesized that the higher capacity realized during the second cycle was principally as a result of a rapid buildup of suction pressure, which did not have sufficient time to dissipate prior to reaching the peak total load of 135 K (601lu'D. A third cycle ofload was also applied in a manner similar to the second cycle, with similar results. It was concluded that the appropriate capacity of the footing, excluding the suction reaction, was 109 K (485 kN) and that it would be reasonable to assign a unit value of suction pressure at first failure equivalent to that which developed at first failure in Footing A (3 psi). Hence, had loading continued at the rate employed in the early stages of the first cycle, a total peak capacity of about 128 K (570 lu"\J") would have been realized. Note that this value is considerably lower than that for Footing A, despite the fact that Footing B had a diameter of 90 in. (2.29 m) compared with 72 in. (1.83 m) for Footing A. The surface deformation patterns and interpreted failure mechanism for Footing B are shown in Figs. 11 and 9, respectively. Less difference in the soil deformation adjacent to the footing and shaft deformation was evident for Footing B than for Footing A. Major surface deformation was also confined to a zone within 60 in. (1.5 m) of the face of the shaft, suggesting a failure body more nearly cylindrical than that for Footing A. The cracking pattern on the surface was similar to that for Footing A, except that only short segments of tangential cracks developed. Footing C The displacement-suction pressure-time relation for Footing C is shown in Fig. 12. Since the intent of the test was to investigate the behavior of the footing under sustained monotonic loading, the results have been plotted as functions of the logarithm of time. It is normally assumed that log-linear displacement-time relationships are indicative of stable behavior. Displacement-time relations for loads up to 160 K (712 kN) (73 percent of the capacity of Footing A) are essentially log linear; with minor variations due to thermal effects. Suction pressures on the order of 1.2 psi (8.3 kPa) or less were developed after first applying each load. Within several minutes these pressures had reduced to 0.4 psi (2.8 kPa) or less and remained essentially constant for the remainder of the load increment. Upon application of the final increment of load, which brought the total load to 190 K (846 kN), a decidedly nonlinear displacement-log time relation was observed, which indicated failure. Suction increased, rather than decreased, with time, in response to the large deformations generated during the maintenance of the final load. However, due to the slow rate of movement, the magnitudes of suction never exceeded 2.3 psi (16 kPa). The total capacity of Footing C, hlded over a long period of time, minus the suction pressure reaction at failure, was 181 K (805 kN), which was almost identical to the total capacity minus suction pressure reaction at large displacements in Footing A. As in the case of Footing A, the pore water suction was essentially identical to the total suction each time readings were taken. No discernable soil surface cracking pattern was evident in the sustained, monotonic loading portion of the test, although when the footing was unloaded and reloaded, a cracking pattern and a soil surface deformation pattern developed that resembled those for Footing A.
122
TRANSMISSION
LINE TOWERS
FOUNDATIONS
Upon unloading and reloading Footing C in a rapid monotonic manner, the load vs. deformation pattern shown in Fig. 13 ensued. The total capacity increased to a total of 219 K (974 kN), but a corresponding suction pressure of 8.8 psi was generated during reloading, which converts to a suction reaction force of 36 K (160 kN), leaving a total force less suction reaction force at failure of 183 K (814 kN), essentially identical to the equivalent capacity measured in the sustained-load portion of the test In Fig. 13 a translated graph of uplift force (T) vs. deflection for the fIrst (virgin) cycle test for Footing A is also shown. Note the almost perfect resemblance to Cycle 2, Footing C. Footin £ D The results of the test on Footing D are summarized in Fig. 14 in the form of displacement versus cycle number for various magnitudes of load amplitude. At the lowest value of load amplitude (70 K (312 kN)), the behavior was essentially elastic to 125 cycles. The behavior at a load amplitude of 120 K (534 kN) appears at fIrst to be erratic. The variable slope of the displacement-cycle number relation is due, however, to a variable cycle period. The steeper slopes correspond to long periods (in the order of 5 to 15 min.), while the flatter slopes correspond to short periods (in the order of 3 min.). The behavior is generally log linear and stable to 100 cycles of applied load. Reduction of the load amplitude to 70 K (312 k!\T) again resulted in elastic behavior. However, abrupt failure was observed after application of 6 cycles at a load amplitude of 160 K (712 kN). Suction pressures generally followed the cyclic trend of the loads. During the fIrst set of cyclic loads at 70 K (312 kN), total and pore water suction values ranged from 1.5 psi (10.3 kPa) during load application to -1.0 psi (-6.9 kPa) during load removal. The negative value of suction (positive total pressure) is probably due to the recompression of air inducted into the total pressure cavity during the loading portion of the cycle and complete return of the base of the footing to zero total displacement during the unloading part of the cycle. Values of suction pressure measured 30 sec after application and removal of load on selected cycles at the failure load amplitude of 160 K (712 kN) are shown in Fig. 15. The total suction pressures, which again were virtually identical to the pore water suction pressures, were generally larger than the suction pressures at corresponding displacements in the monotonic tests. Here, the suction remained positive even during unloading. The maximum load minus corresponding suction pressure (11.3 psi (77.9 kPa)) reacting over the base of the footing is only 114 K (507 kN), compared to about 180 K (800 kN) for the rapid and sustained monotonic loading on Footings A and C, which were of comparable dimensions to Footing D. This observation suggests that cyclic loads of increasing amplitude had a severe degrading effect on the maximum frontal soil resistance (Q) available above the bell.
QUMrrIFICA TION OF OBSERVED CAPACITIES The salient results of the tests are summarized in Table 1. Based on the values reported in that table and on the observed phenomena described in the preceding section, it is possible to develop a simple, coherent, phenomenologically-based equation for
DPUFf IN JOINTED CLAY
-' Q) Q C
2 453 (/) .c ~ () (f) E D01 C ~ 0
123
• T = 70K oT=120K
6 T = 70K (Reload) c T = 160K
x - 3 min. period y - 5-15 min. period
10
100
Number of Cycles
1,000 (N) Footing
Fig. 14. Displacement Vs. Cycle Number, Footing D (1 in. = 25.4 mm; 1 K = 4.45 kN)
-- --
Loaded
..•....
c:
,
";n
o D';:: -() ::J
I
Q) •••• ::J
(f)
(/) CO -
(/) Q)
o •...
---
f-D.
/
Footing Unloaded Period::: 6 min
2
3
Number of Cycles
456 (N)
Fig. 15. Total Suction Vs. N, Footing 0, 160K Load (1 K = 4.45 kN; 1 psi = 6.89 kPa)
I
8 7
/~From
6 5 ::J
Z
3
/
2 1
o
o
/ /
/
/
for Flat Plates
/l· /
4
//
Breakout Theory
(This study)
,..--
__ ..-
..---
(rp = 0)
(9)
__
..- ..---~ in-situ tests on belled Nu = 4.64 From «D/B)-0.77) footings in fissured cia y (1)
2
3
D/B
Fig. 16. Factor Nu Vs. DIB
Sat 46 78 9.4 2.5 2.2 1.5 1Cyclic 47 12 36 72-S(in.) 7.4 124 Movement 1(residual 60 10.2 19 36 48 219 6Condition 145 128 9.4 9.7 220 1.69 Sustained TRANSMISSION (inferred) (K) (K) = 143) TuLINE TOWERS FOUNDATIONS 1.56 (3.0 psi) 0.99 Rapid (K) (8.8 (2.3 190 1.61 u (wf) Ab -TPeak:Load yDAb Reload 1 psi = 6.89 kPa) (K) D Rapid Table 1. Summary of Salient Results (1 K = 4.45 kN; 1 in. = 25.4 mm; 1 ft = 0.305 m; epth DIB Loading "fD
Table 2. Comparisions of Dead-Weight-of-Cone Capacity with Measured Capacity (1 K = 4.45 kN) Footing
T u -S (K)
Capacity Computed from Dead Weight of Truncated Cone (K)
A
208
132
B
109
101
C
181
152
D
114
125
UPUFf
IN JOINTED CLAY
125
describing the peak, rapid monotonic, failure loads for the test shafts. Assuming that undrained failure occurs in the day, Eq. 1 can be rewritten in the form developed for breakout of flat disks near the soil surface ( 9 ), which has been shown to model accurately the capacity of the top surface of helical anchors in homogeneous clay ( 5 ):
(2)
where
Nu Su
= 4.64 ( DIB - 0.77 ) (using the triaxial data), = 4.35 (DIB - 0.77 ) (using the CPT data); = average undrained shear strength from the base of the footing to a level
2 ft (0.61 m) below the ground surface from either UU triaxial shear strength profile or CPT shear strength profile (Nk = 19); 'Y
D ps
= soil/concrete unit weight; = footing depth; = maximum total suction pressure at failure at the base of the footing
(approximately 3 psi (20.7 kPa)); Au
=
Ab
= 1tB2j4;
1t (
B2 - b2) j 4, where B = bell diameter and b = shaft diameter; and
All factors in Eq. 2 were measured directly, except for Nu' which was then calculated from Tu (Table 1) and fitted linearly to DIB. Eq. 2 can be modified to account for the observed effects of sustained and one-way cyclic loading by including two factors <1>1and <1>2'as described in Eq. 2a: (2a)
where
<1>1 = shear strength degradation factor = 1.0 for rapid monotonic loading, = 0.85 for slow (sustained) monotonic loading, = 0.45 for progressively increasing one-way cyclic loading; <1>2 = suction factor = 1.0 for loads applied for less than one minute, = 0.1 for loads applied for longer than five minutes (for small displacements).
The N u factors are shown as functions of DIB on Fig. 16, on which are also plotted results from model and full-scale tests in fissured clay ( 1 ) and a theoretical relation for flat circular disks in homogeneous soil. The factor for the smaller DIB in the present study is near the corresponding value from Ref. 1, which suggests that the low capacity of the shallowest footing (Footing B, DIB = 1.0) was associated with opening of joints during loading, which is not reflected in the shear strength measurements. The factor for DIB = 1.67 is much closer to the theoretical relation for surface breakout of flat disks,
126
TRANSMISSION LINE TOWERS FOUNDA nONS
suggesting less effect from the opening of soil joints. Another common method of computing uplift capacity of shallow belled footings is the dead-weight-of-cone method. The capacity of the footiing is taken to be the dead weight of the footing plus the soil inside a truncated cone that rises from the perimeter of the base of the footing and makes an angle of 30 degrees with the vertical. This proposed failure block is not dissimilar to the inferred failure block for Footing A. No shearing resistance is assigned to the soil, and zero suction is assumed. Table 2 summarizes the results from this method and compares them to the measured capacities minus suction resistance. The dead-weight-of-cone method predicted capacities that were conservative for the monotonic tests, although the error was small for DIB = 1.0 (Footing B), and predicted a capacity that was somewhat too high for the cyclic test (Footing D).
CONCLUSIONS
The following conclusions are drawn from this study: 1. The uplift capacities of shallow belled footings in jointed Beaumont Clay were influenced by surface effects and the presence of joints in the soil, but more predominantly at DIB = 1.0 than at DIB = 1.67. The footing tested at D/B = 1.0 apparently had very low frontal soil resistance above the bell, which implies that such shallow embedment was ineffective in the jointed soils at the test site. 2. Suction (primarily pore water suction) contributed significantly to short-term uplift capacity, although large displacements were necessary to affect total suction pressures approaching 1 atmosphere. Suction also was found to dissipate rapidly after application of an increment of load but not to disappear entirely under sustained loads. These characteristics can possibly be considered in design in the Beaumont Clay if the duration of applied loads is known. 3. Cyclic loading produced a severe loss of frontal resistance in the soil above the bell, while sustained loading produced only minor soil capacity reduction. 4. The capacities of the test footings are expressed in simple mathematical form in Eqs. 2 and 2a. These equations are rational, although they contain empirically evaluated terms, and incorporate the most important characteristics of the test-footing/soil system. They are not proposed for general design use.
APPENDIX - REFERENCES 1. Adams, 1. I., and Radhakrishna, H. S., "Uplift Resistance of Augered Footings in Fissured Clay," Canadian Geotechnical Journal, Vol. 8, 1971, pp. 452-462. 2. Bonar, A, J., "Uplift Resistance of Tower Foundations," Research Report to Houston Li~hting and Power Company, Department of Civil Engineering, University of Houston, August, 1961.
UPLIFT IN JOINTED
CLAY
127
3. Kulhawy, F. H., "Uplift Resistance of Shallow Soil Anchors - An Overview," Uplift Behavior of Anchor Foundations in Soil, Ed. by S. P. Clemence, ASCE Special Technical Publication, Oct. 1985. 4. Mahar, L. J., and O'Neill, M. W., "Geotechnical Characterization of Desiccated Clay," Journal of Geotechnical Engineering, ASCE, Vol. 109, No.1, Jan. 1983, pp.
56- 71.
5. Mooney, J. S., Adamczak, S., Jr., and Clemence, S. P., "Uplift Capacity of Helical Anchors in Clay and Silt," Uplift Behavior of Anchor Foundations in Soil, Ed. by S. P. Clemence, ASCE Special Technical Publication, Oct. 1985. 6. O'Neill, M. W., and Sheikh, S. A., "Geotechnical Behavior of Underreams in Pleistocene Clay," Drilled Piers and Caissons II, Ed. by C. N. Baker, Jr., ASCE Special Technical Publication, May 1985. 7. Sheikh, S. A., O'Neill, M. W., and Yazdanbod, A., "Uplift Behavior of Shallow, Fun-Sized Underreamed Footings in Beaumont Clay," ReDon No. UHCE 86-5, Department of Civil Engineering, University of Houston - University Park, June, 1986. 8. Turner, E. A., "Uplift Resistance of Transmission Tower Foundations," Preprint, ASCE National Convention, Houston, Texas, Feb. 1962. 9. Vesic, A. S., "Breakout Resistance of Objects Embedded in Ocean Bottom," Journal of the Soil Mechanics and Foundations Division, ASCE, VoL 97, No. SM9, September, 1971, pp. 1183 - 1205. ACKNOWLEDGMENTS The authors thank Houston Lighting and Power Company for sponsoring this study, for providing construction personnel and for permitting publication of the results. They are also grateful for the assistance and technical support given by Dywidag Systems International, USA, Inc., Farmer Foundation Company, and McClelland Engineers, Inc. The participation of several staff members and present and former students at the University of Houston - University Park in the performance of the field tests, especially David Menzies, Ketan Kapasi, Dennis Paul, Harry Yearsley, Todd Dunnavant and Brad Gana, is also acknowledged.
UPLIFT CAPACITY OF DRILLED PIERS IN DESERT SOILS A CASE HISTORY By Byron Konstantinidis1,
Albert].
Pacal2,
Arthur W. Shivel/
ABSTRACT
This paper presents an evaluation of uplift capacity of drilled piers in desert soils based on comprehensive geotechnical investigations and fullscale load tests performed at four sites. The geotechnical investigations included borings, laboratory tests, pressuremeter tests, and cone penetration tests. The soils at the four sites ranged from stiff clays to dense gravelly sands. The paper includes a comparative evaluation of state-of-the-art uplift capacity prediction methods available to the geotechnical engineers.
INTRODUCTION
Drilled cast-in-place piers are the most common foundation type used for high-voltage transmission line towers located in the deserts of the Western United States. The lattice type towers commonly used for such transmission lines are typically supported on four piers. Due to high overturning loads imposed by wind loads or line tension, the design of these piers is generally governed by uplift capacity considerations. Typical design uplift loads for high voltage (230 KY and higher) lines are on the order of 100 kips (445 kN). However, at angle (corner) towers, design uplift loads can exceed 300 kips (1335 kN) for sustained line loads and 500 kips (2225 kN) for transient line loads.
GEOFON, Inc., Cypress, California. ?lYice President, -Civil Engineer, Department of Water and Power, Los Angeles, CA. 3Manager of Quality Assurance, Dept. of \Vater and Power, L.A., CA.
128
UPLIFT CAPACITY OF DRILLED PIERS
129
This paper presents an evaluation of uplift capacities of drilled piers in typical desert soils based on comprehensive geotechnical investigations and full-scale load tests performed at four sites. The main purpose of these tests was to confirm foundation designs for a major 500KV transmission line in the Southwestern United States. The foundation designs were based on empirical in-house techniques that generally resulted in more economical designs than those indicated by conventional analytical methods. A secondary purpose of these tests was to evaluate the accuracy of stateof-the-art methods based on in-situ soil exploration methods in predicting the uplift behavior of drilled piers in desert soils. Geotechnical investigation at all four sites included cone penetration tests, borings, and laboratory tests. Pressuremeter tests were also performed at three of the four sites (the soils at the fourth site were too coarse for such tests).
SITE
CONDITIONS
A detailed geologic reconnaissance was completed along the entire transmission line alignment before undertaking the foundation studies described in this paper. Based on the results of this reconnaissance, four sites, representing the range of soil conditions present along the alignment, were selected for detailed studies. The range of soil conditions at these sites is representative of desert soils, in general. Characteristically, desert soils are overconsolidated by desiccation, exhibit some cementation, and have relatively high shear strength. The subsurface conditions at the four test sites are summarized below. SITE
NO.1
- DELTA
Site No.1 is located in the Sevier Desert, 20 miles (32 km) southwest of Delta, Utah. This area was once part of Lake Bonneville. Soils in the upper 19 feet (5.8 m) at the site consist of silty clays of low to medium plasticity. The consistency of these soils ranged from firm to very stiff. For the foundation evaluations presented herein, the Delta site is considered a "stiff clay" site. Medium dense to dense silty sands underlay the clays. Groundwater was encountered at a depth of 18 feet (5.5 m). The Moisture content in the soils above the groundwater table was variable, ranging from slightly above the plastic limit to slightly below the liquid limit.
130
SITE
TRANSMISSION
NO.2
LINE TOWERS
FOUNDATIONS
- CALIENTE
Site No. 2 is located on a very extensive alluvial fan in the Delmar Valley, 20 miles (32 km) southwest of Caliente, Nevada. Soils at this site consist of dense, slightly to heavily cemented silty and gravelly sands with occasional cobbles. The moisture content of these soils was generally very low. No groundwater was encountered in the borings. SITE
NO. 3 - ALAMO
Site No. 3 is located within the Delmar Dry Lake, 16 miles (26 km) east of Alamo, Nevada. The dry lake is located at the bottom of Delmar Valley, at the base of very long alluvial fans. Soils at this site consist of very stiff to hard silty clays. The in situ moisture content of these soils was near the plastic limit. SITE
NO.4
- BAKER
Site No. 4 is located in the Silurian Valley, 10 miles (16 km) north of Baker, California. This site is located on a relatively short alluvial fan. Soils at this site consist of medium dense to dense silty sand and gravel. The moisture content of these soils was very low (less than 2 percent). No groundwater was encountered in the borings. Results of in-situ tests performed at the four sites are presented in Figure 1. The cone penetration tests were performed using a truck mounted electric cone penetrometer with a maximum thrust capacity of 20 tons (178 kN). Pressuremeter tests were performed using a TEXAM pressuremeter (1) in pre-drilled small diameter boreholes. At Site No. 1 the boring was drilled with a hand auger. At Site No. 2 rotary wash with foam and mud was used. At Site No. 3 rotary drilling with airinjection was used. At each location tests were performed at four depth intervals. The coefficient of earth pressure at rest (Ko) was obtained using a new method (1) which is analogous to the determination of preconsolidation pressure from laboratory consolidation tests. Geotechnical parameters summarized in Table 1. obtained by three-point moisture conditions.
derived from field and laboratory tests are It should be noted that shear strength data were direct shear tests performed under in situ
UPLIFf CAPACITY OF DRILLED PIERS
CONE FRICTION
PENETRATION
RE£lnANCE 10
15
TSF (KG/CM2) 100
I
o
PRESSUREMETER DATA
DATA
CONE RESISTANCE
TSF(KG/CM2) 20
TEST
200
300
400
FRICTION (~;) RATIO 500
131
1----
TEST
POH (k Pal PL· (102kP.)
600 0
60
80 0
8
2
o en a::
>5
u.J
u.J
>-
u.J
u. ~
u.J
::E
10
<; ,... 0.. u.J
a
4 15
:r
>Q. u.J
a
20
SITE NO.1
DELTA
en >u.J
c:
5
u.J
>-
u.J
u. ~
u.J
::E
10
~
,...
0.. 15
>=
',,"
Q.
a
u.J
a
20
SITE NO.2
CALIENTE o en a::
..:,
I'
u.J
>w ::E
~ 4
:t
a
15
W
a
20
SITE NO.3
ALAMO i!
I
!i
I
I
I
!
I
u.
:
,
I
1-0 S5~110 ~ 0..
,I I' ',~
,', i
UJ
a
:r
>Q.
:
15·
'I
"I
I
ND TEST
en
I
a::
2
w >-
w ::E
z 4 ,
20 II
I
i
I
I PERFORMED I
;
SITE NO.4
,
~LJ I
I
,~O TEST
; PERFORMED
BAKER
FIGURE 1: IN-SITU TEST RESULTS
~ >= 0.. W
a
TRANSMISSION LINE TOWERS FOUNDATIONS
132
TABLE 1 - LABORATORY TEST RESULTS
l
uf A Y
(kPa) DESCRIPTION
,
-- ----~- I
-- -
LABOR A TORY TEST DATA
, 95 , 11.521 LL 99 11.591 19 41 13 42 25 27 32 28 16 44(;;;cm3) 17(~.) PL 0.58 0.40 15610.58 1381 1561 Q I,IIII IIIpcfI26 W 16 99 8211.311 11.591 23 28 0.3517 1341 C I 11511.841 Yd
SOIL
I
U)
t;; w ...
5 2
~
~
~O
:I:
4
•...
o
:I: •... "-
15
fu
c: w t;; :!:
w
o 20
SITE NO.1 DELTA
-•
114 NP 11.831 NP NP 43 III I 48 iNP 11311.811 NP GRA VELL 10711.711 11011.761 Y SAND I occ ••. on.1
cobble1
w
I I
•... ... 105 w
•...
~ 0 ~
1S 20
--
00
i
0
U)
c: w •... w :!:
I
II
~ 4
:I: •... "-
w
o
SITE NO.2
CALIENTE
o 0.501481
t;; w ...
5
•...
o
w
0~ .w.. "-
6
0 15 w
20 U) 31II 14
~I
•...
SANDY CLA Y ICLI sandi.r "'ltho~th
becoming
15
SITE NO.3
•...;:::
53 40
SIL TY TO
~ 10 :I: ~
I
10 15 0 20
ALAMO
I
NP 114 11.831 47 41 NP !i 11611.861 11811.891 NP NP 34 21 0.151141 38 NP I 10e 11.73) 0.101101 NP ISP·SMI 0.30 1291
'" c: w 2
••..
w :!:
Z :I: •... "-
W
o
SITE NO.4
BAKER
UPLIFr
TEST
CAPACITY
OF DRILLED
PIERS
133
PIER CONSTRUCTION
At each site four test piers were constructed although only two or three were tested in uplift. Test piers were constructed using a standard truck mounted bucket auger rig with. Nominal design dimensions of the test piers were 24 (61 cm) inches in diameter and 10 or 15 feet (3.0 or 4.6) in length. However, as-built dimensions differed slightly from design dimensions, as shown in Table 2. The tops of the test piers were recessed approximately 6 inches (15 cm) below grade, and the top 6 inches (15 cm) of the piers were formed with sona-tube. For analysis purposes the load carrying portion of the pier was assumed to begin approximately 1 foot (30 cm) below grade. Test piers were reinforced with six No. 6 longitudinal bars and a spiral of No. 3 bar, all bars conforming to ASTM grade 40. A 6 x 6 x 3/8 :\.36 steel angle with bolted on c1eets was embedded 3.5 feet (1 m) into the pier. The centroid of the angle was positioned over the center of the pier . Vertical load from the test rig to the pier was transferred via the steel angle described above.
TABLE 2 - "AS BUILT" PIER DIMENSIONS NO. 431 10.5 TOTAL PIER DEPTH DIAMETER lower 2(cm) 2.7 feet feet (meters) 10.1 10.0 15.0 8.3 (3.2) (3.1) (3.0) (4.6) (2.5) inches 15.5 (4.7) 10.5 (3.2) 5.6 upper feet 26 ~1/2 (66) 26 27 25 ~COMMENT (66) (69) ~ 8.5 (67) (65) SITE 21 feet 36~(91) 36 ~ (91)
134
TRANSMISSION LINE TOWERS FOUNDATIONS
LOAD TESTING
Uplift loads were applied to the test piers via a 300 kip (1335 KN) hydraulic jack mounted between two W27 x 84 wide flanged beams. A distance of 10 feet (3 m) was maintained between the pier and points of support for the test rig. Load increments were maintained for a minimum of 5 minutes to a maximum of 60 minutes. Equilibrium was usually established in about 15 minutes. Due to system limitations, it was very difficult to maintain loads for extended periods of time. Pier deflections were monitored utilizing two independently supported dial indicators mounted 180 degrees apart relative to the top of pier. Dial indicators were graduated to 0.001 inch (0.025 mm). In addition to monitoring pier deflections, ground disturbance was also monitored. Hubs were driven into the ground radiating out along a tangent to the pier. A fine piano wire was stretched taut directly above the hubs. Ground deflections were measured with a scale graduated to 0.01 inch (0.25 mm). Load-deflection curves are presented in Figure 2. It should be noted that the maximum applied load was limited to 200 kips (890 K1'\', 2/3 of the jack capacity). In all cases a "yield" load could be detected within this range although in three cases (Pier No. 1 at Site No. 1 and both piers at Site No.3) the ultimate capacity appeared to be greater than 200 kips (890 KN). Ground heave measurements around the test piers indicated movement in all cases. The heave adjacent to the pier approached the vertical deflection of the pier gradually diminishing with radial distance. The maximum radial distances where ground displacement was detected (or projected based on adjacent monitoring points) are presented in Table 3.
~
c.. ...J -4 SITE NO.1 -2 -.2 -.3 -.3 a --6 6a-.1 SITE NO.3 -2I-oj SITE NO.2 DEFLECTION IN BAKER CALIENTE -.1 III-.2 800 DEFLECTION IN INCHES MILLIMETERS :;:) :J 135 UPLIFTRESULTS CAPACITY OF600DRILLED PIERS LOAD TEST I I --"1
120 <:
c.. ...J
120 80 160 ~I 80 160 40
.:JI!.
--
URE 2:
TABLE 3 - APPROXIMATE LIMITS OF GROUNDREAVE
413214PIER 3
NO. DIAMETER OF HEAVED feet (meters) 10.5 15.5 (3.2) (4.7) 89feet (2.4) 10.1 (3.1) 10.0 (3.0) (2.7) 8.3 (2.5) 11 10 (3.4) (3.0) 15.0 (4.6) 11 (3.4) (meters) LENGTH
GROUND OF PIER
:i0
200 ...J a200 Z 600 SITE NO.4800 400 400
~ 0 Z ~
uc.. u....
.:JI!.
~
136
TRANSMISSION LINE TOWERS FOUNDATIONS
EVALUATION
OF RESULTS
Prior to conducting the full scale load tests, uplift capacities were predicted using two techniques, in this paper briefly referred to as the EPRI method and the CPT method, respectively. Ultimate uplift capacities were also predicted by others (2) based on pressuremeter data. The results are summarized by Briaud et. al. (3) and will not be repeated here. The EPRI method is based on shear .strength data obtained from laboratory tests and in situ lateral stress estimates either obtained by pressuremeter tests or analytically derived from the geologic load history of the site. The method is described in Chapters 8 and 9 of a comprehensive state-of-the-art research report (4) prepared by Cornell University for the Electric Power Research Institute. The method is based on the assumption that failure occurs primarily as a cylindrical shear surface along the perimeter of the pier. The uplift capacity of a straight-shaft assumed to consist primarily of side friction andlor adhesion along the cylindrical surface of the pier plus the weight of the pier. Under drained condition, the ultimate uplift capacity derived from side friction, F s , is
F s=
iT
B
-
"'(z
K(z)
tan 0
dz
Where
B
IS the diameter
D
IS the total embedded length of the pIer IS the average
of the pier
effective
unit weight of the soil
z and dz
are the average
K(z)
is the coefficient of horizontal stress (assumed to be equal to the at rest coefficient,
depth and thickness
of a layer
Ko )
is the angle of friction for the soil to concrete interface (equal to the soil friction angle, , for cast-in-place piers) For undrained loading conditions, typically used for saturated, fine grained soils under quick loading conditions, the uplift capacity due to adhesion is expressed as
F s=
iT
B
S u (z)
dz
UPLIFT CAPACITY OF DRILLED PIERS
137
where
Su (z) a(z)
IS
the undrained
shear
strength of the soil in a given layer
is an empirical adhesion factor, for concrete piles ranging between 0.4 for hard soils to over 1.0 for soft soils
For short piers in hard soils, it is also recognized that a composite failure surface consisting of a cone near the ground surface and cylindrical surface at depth can develop resulting in somewhat lower capacity than would be predicted by the cylindrical failure surface model. The CPT method is based on side friction values obtained from electric cone penetration tests. This method is described by Schmertmann (5) based partly on research performed by Nottingham for driven piles. This method is also based on the assumption that failure occurs along a cylindrical surface. However, for granular soils reduction factors are applied to side friction values near the ground surface to account for reduced confinement effects. For granular soils, the ultimate side friction resistance, F s . for piles in compression is
F s=
Ks
z
88 ~ o ~
88
fs
A
s
L ) 88""
+
fs
A
s
Where Ks
is ratio of unit pile friction to unit sleeve friction (for short concrete piles, it ranges between 1.2 and 1.5)
Fs
is unit sleeve friction penetrometers
A
pile-soil contact considered
s
For cohesive
soils,
side friction
resistance
area
for
is estimated
from CPT's
the
by
Where a
IS
the adhesion factor,
defined earlier
depth
using electric
interval
being
138
TRANSMISSION LINE TOWERS FOUNDA nONS
For piles in tension, Schmertmann (5) recommends that the frictional capacities be reduced to 2/3 of the values computed by the above formulae. For drilled piers, Schmertmann (5) recommends a further reduction to 3/4 of the computed values. Thus, with the combined reduction factors, the frictional capacity of drilled piers in uplift is expected to be equal to half of the frictional capacity of driven piles in compreSSIOn. Upon completion of the load tests, the predicted uplift capacities were compared with the actual observed capacities. The results of these evaluations are summarized in Table 4. It should be noted that the actual capacities were defined by four different methods described in the referenced EPRI publication (4). In some cases the range of capacities indicated by the four methods was very wide making objective comparisons difficult. In all cases, however, the capacities indicated by these methods were lower than the ultimate uplift capacity. Thus, they are more representative of a "yield" or "plunging" load rather than ultimate capacity. In practice the writers have found the "slope tangent" method to be most convenient to use for evaluating uplift capacities of drilled piers in desert soils. In addition to its simplicity, this method consistently produced capacities slightly below the ultimate measured capacities, at tolerable measured deflections (less than 1/4 inch or 6 mm).
TABLE 4 - PREDICTED AND ACTUAL
-
160 80 144 1 142 145 122 91 115 >142 175 120 140 76 85 80 62 79 157 180 160 145 Pier 3 No. >200 88 137 111 82 126 >159 >160 210 88 78 Intersect 95 178 90 110 118 130 180 85 135 90% 75 EPRI 76 90 130 78 125 105 9 0 400 145 CPT 90 75 203 Avg. Tangent Tangent Log-log PREDICTED
CAPACITIES
UPLIFT
ACTUAL CAPACITIES Slope (KIPS)
(KIPS)
CAPACITIES
UPLIFT CAPACITY OF DRILLED PIERS
139
The EPRI method predicted capacities that were typically slightly lower than the actual capacities indicated by at least three of the four methods. In general, the capacities predicted by the EPRI method were within 30 percent of actual capacities. It should be noted that the predictions made using the EPRI method were based on very high Ko values indicated by the pressuremeter tests (see Figure 1). Had the lateral pressure coefficient, K, been limited to a value less than 1.0, as it is commonly done in western geotechnical practice, the predicted capacities would have been less than half of the actual values. The CPT method predicted the uplift load capacities at three of sites even more accurately than the EPRI method. However, at 3 (Alamo) it predicted uplift capacities that were substantially than the actual capacities indicated by at least three or the four This discrepancy can be attributed to a combination of two factors; namely
the four Site No. greater methods. potential
a)
As can be seen in the load-deformation curves (see Figure 2), the actual ultimate uplift capacity for both piers at this site was much greater than the "yield" load indicated by three of the four techniques, and quite likely much greater than the 200 kip load limit of the test. Thus the discrepancy may not be as significant as it appears.
b)
The site soils were highly desiccated and possibly had planes of weakness that would have reduced the uplift capacity. Such secondary structure would be not detected by cone penetration testing.
CONCLUSIONS
Based on the results of the geotechnical investigations described in this paper, the following are concluded:
and
load tests
1.
The EPRI method consistently provided reasonably conservative uplift capacity estimates despite the liberal earth pressure coefficients used in the analyses.
2.
The CPT method generally provided very realistic estimates of the ultimate uplift capacity with the possible exception of Site No. 3 where the method overestimated the capacity of at least one pier. The results of these studies are encouraging considering the fact that cone penetrometer testing is one of the more economical subsurface exploration methods and that adequate penetration was achieved even in dense gravelly and
TRANSMISSION LINE TOWERS FOUNDATIONS
140
cobbly sands. However, the data base for drilled applications is very limited and more research is needed the effect of limiting conditions such as secondary structure (fissures, cracks, bedding planes, cementation, and moisture content variations. 3.
The mode of failure in all cases formation of a conical surface. failure did not appear to adversely uplift load predictions.
pIer into soil etc.)
appeared to involve the However, this mode of affect the accuracy of
ACKNOWLEDGMENTS
The studies described in this paper were sponsored by the Intermountain Power Agency, as part of the design effort for the Intermountain Power Project. The studies were coordinated by the engineering staff of the Los Angeles Department of Water and Power, who also performed the load tests. Subsurface exploration and laboratory testing services were provided by the Earth Technology Corporation. Pressuremeter testing services were provided by Briaud Engineers.
APPENDIX I - REFERENCES
1.
Briaud, ].-L., Pressuremeter"
Babb, L., Capelle, J .-F., "The TEXAM Geotechnical testing Journal ASTM 1983.
2.
Briaud Engineers, "Foundation Testing for Electric Power California", unpublished report,
3.
Briaud, ].-L., Pacal, A. ]., and Shively, A. W., "Power Line Foundation Design", Proceedings of International Conference on Case Histories in Geotechnical Engineering, Saint Louis, May 1984.
4.
Cornell University, "Transmission Line Structure Foundations for Uplift-Compression Loadings", Electric Power Research Institute Publication, EPRI EL-2870, Research Project 1493-1, February 1983.
5.
Schmertmann,]. H., "Guidelines for Cone Penetration Test, Performance and Design", U. S. Department of Transportation, publication FHW A- TS-78-209, May 1978.
Investigation by Pressuremeter Line in Utah, Nevada, and July 1982.
UPLIFT CAPACITY J
OF DRILLED PIERS
141
APPENDIX II - NOT ATION
, The following
I I
i
ss dz cF D
symbols
were used in this paper:
A
Cohesion Unit friction from data Ultimate Pier side diameter side friction resistance layer thickness Total embedded length ofCPT pierarea Incremental soil-pile contact
K, K(z)
Coefficient
of horizontal
stress
at failure
Ko
Coefficient
of horizontal
stress
at rest
Ks
Ratio of unit pile friction
LL
Liquid Limit
PL
Net limit pressure
P OH
Horizontal
PL
Plastic
Limit
w
Water
content
z
Depth below ground surface
a
Adhesion factor
'Y
Average
B f
from pressuremeter
soil pressure
effective
'Y d
Dry unit weight
o
Angle of friction
¢
Friction
to unit CPT sleeve friction
(also
test
at rest from pressuremeter
known as Tomlinson's
unit weight
between soil and concrete
angle of soil
factor)
test
UPLIFT CAPACITY OF DRILLED DRIVEN PILES IN GRANULAR Keith
ABSTRACT:
Southern
D. Tucker*,
SHAFTS AND MATERIALS
A. M.
California
ASCE
Edison
Company
has
performed field uplift load tests on cast-in-place drilled shafts and driven piles along transmission line routes and generating facilities within its service territory. Field exploratory borings and cone penetrometer test soundings were placed at many of the test sites to identify the soil types, densities and strength characteristics of the subsurface materials. In this paper, results from 91 field uplift load tests are utilized to evaluate design methodologies for computation of ultimate uplift capacities. The field load-deflection results are normalized to predict behavior of the drilled shafts and driven piles. Correlations of side friction factors with shear strength and foundation geometry are given for use in predicting the uplift capacity of foundations in granular materials.
INTROOUCTI
ON
The Southern California Edison Company (SCE) has performed more than 100 field uplift load tests on drilled shafts and driven piles over the past 50 years for transmission line structures throughout the SCE service territory. These tests provide a large data base to evaluate design methodologies for estimating the ultimate uplift capacity and associated deflections of drilled shafts and driven pile foundations in granular materials. LOCATION
OF
FIELD
The field uplift line routes from
LOAD
TESTS
load tests were conducted along eight 1936 to 1985 and at six SCE facilities
transmission from 1941 to
1986. These test locations ranged from the Tehachapi mountains southeast of Bakersfield, California, to coastal sites near Ventura, California, and as far east as the Colorado River at Blythe, California. The SCE service territory and location of the field load tests are shown in Figure 1. *Geotechnical
Engineer,
Southern
California
142
Edison,
Rosemead,
CA
DRILLED
SHAFTS
AND DRIVEN
143
PILES
FRESNO
o
BAKERSFIELD
o MAGUNOENPASTORIA T IL
~\ » r-1:tJ =T\IN
0'0 :::>\' ,,>
ORMONO BEACH GEN. STATION
OEVERS-PALO
i
:;
SCE REPORT NO. 124 SITES ~VEROE PALM ~ SPRINGS ,0
FIGURE 1 SOIL
TIL
NEWPORT BEACH
)
.' /
LOCATION OF SCE UPLIFT LOAD TESTS
CONDITIONS
The soil conditions encountered along the transmission line routes ranged from 'Wind-blo'Wn sands in desert regions to alluvial deposits of dense sands and gravels near mountains. Fractured and slightly 'Weathered sandstones, siltstones and granitic materials 'Were prevalent in the Tehachapi mountain range. The coastal sites in the Los Angeles basin and 10'W-lying areas near the Colorado River consisted of intermixed sand, silt and clay materials 'With ground'Water depths from 2 to 15 feet (0.6 to 4.6 meters). FIELD
EXPLORATION AND LABORATORY DATA
The earlier load tests from 1940 to 1950 'Were performed at sites 'Where a minimum of subsurface information 'Was available. A description of the soil type, consistency and drilling procedures 'Were the primary data obtained in field explorations. From 1950 to 1986, exploratory borings 'Were often placed near the test piles 'With Standard Penetration Tests (SPT) performed to obtain blo'Wcounts at different depths. T'Wo types of samplers 'Were used in the field, the standard split barrel sampler 'With a 2 inch (5.1 cm) 0.0. for SPT tests and a ring sampler 'With 4 inch (10.2 cm) 0.0. to collect relatively undisturbed samples.
TRANSMISSION LINE TOWERS FOUNDATIONS
144
Since 1981, electric Cone Penetration Test (CPT) soundings have also been performed to obtain in-situ strength parameters. A standard electric cone was pushed at a rate of 0.8 in/sec (2 cm/sec) using a 20 ton (89 KN) reaction truck. Both side friction and tip resistance profiles were recorded continuously and used in computing the friction ratios. The laboratory testing program on selected samples consisted of moisture content, unit weight, gradation, Atterberg limits and drained direct shear tests on saturated samples at various consolidation pressures. These soil parameters from laboratory tests are given in unpublished SCE reports and were used in evaluating the load test results. FIELD
UPLIFT
LOAD
TEST
PROCEDURES
Uplift load tests were performed using various equipment and methods. From 1936 to 1980, a steel beam was placed across reaction piles with a hydraulic jack resting on the beam. The load was applied manually and recorded from a pressure gauge attached to the pump. Originally, proof tests were conducted to at least 150 percent of design load with typical vertical deflections of less than 0.15 inch (0.4 cm). The load was then rebounded to zero and the permanent deflection noted. In 1981, SCE fabricated a portable steel tripod test frame which is 10 feet (3.0 m) high and has three legs spaced 18 feet (5.5 m) apart at 120 degree angles from each other. A double-acting hollow plunger hydraulic jack with 150 ton (1335 kN) capacity and 8 inch (20 cm) stroke was used to apply the tensile loads. A 1.375 inch (3.5 cm) diameter, high-strength Dywidag bar extends through the jack and was attached to the top of the foundation. Load tests were conducted by applying a tensile load to the Dywidag bar in increments of approximately 25 percent of design load. The load was typically rebounded to zero from 25, 50 and 75 percent of the design load, then the load was re-applied until the peak value was reached prior to a final rebound. Deflections at the top of the pile foundations were measured using two or more dial gauges with an accuracy of at least 0.001 inches (0.0025 cm). The, dial gauge readings were averaged to obtain the actual vertical displacement of the foundation. BASIC
CONSIDERATIONS
In principle, the soils is shown in vertical equilibrium
uplift capacity Fig. 2a and may equation:
of be
drilled computed
shafts in granular from the following
(1)
DRILLED SHAFfS AND DRIVEN PILES
I f
with Ou resistance depending granular tension
145
uplift capacity, W = foundation weight, Os and Ot = tip resistance. The side resistance on the shearing surface and shearing resistance
=
materials. and suction
The tip stresses
resistance can at the bottom
be of
=
side varies of the
developed from the foundation.
During drained loading, suction is not present and tip tension is normally very low for cast-in-place concrete drilled shafts (5). Since the tensile strength of granular soils is usually low, the tip resistance for the drilled shafts and driven piles was assumed to be zero.
OU
I"
~I
I..~I••
~ Otu
1
0tu 0su A) FORCE
FIGURE 2
The
side
resistance,
B) SIDE AND TIP RESISTANCE
DIAGRAM
Os.
DRILLED SHAFT IN UPLIFT
is
shown
in
Fig.
2b
and
may
be
expressed
as:
( 2) where As = surface area of soil-shaft interface, fs = average skin friction along soil-shaft interface and D = embedded depth of f 0 u n d at ion. The sid ere s i s tan c e va rie sin a par a b 0 1 ic ma n n era 1 0 ng the shaft to a minimum value at the tip of the shaft (7,10). INTERPRETATION
OF
FIELD
LOAD
TEST
DATA
Based on recent SCE structural analysis of transmission line towers, a one inch differential deflection of the tower foundations were considered acceptable for design using ultimate uplift loads. For field load tests where the peak uplift resistance occurred at displacements greater than one inch, the ultimate uplift capacity
EFLECTION
1:/
TRANSMISSION LINE TOWERS FOUNDATIONS
146
was established as the applied load at a vertical deflection to one inch. Typical applied load versus vertical deflection from field uplift load tests are shown in Figure 3 for shafts and driven piles.
20
!/~I ~~----1
/ /
II I I /
0 I -- III--r;
Cl. 100 ::J Cl. 1.2 0.2 0.6 0.4 0.8 1.0 140 0 40 60 PIER /I OF AT 78 KIPS <.)AT 80 40 J 101 KIPS 80 140 KIP~ __ I I DC/1I •••.I." I ("\~n""1~n 1 INCH I /' I I / I PEAK LOAD=120LOAD KIPS OF I/ULTIMATE
"1//<~I
FIGURE 3
>=>
c;,
equal curves drilled
Cl. ">
DISPLACEMENT B) DRIVEN CONCRETE
BELLED ;/ ULTIMATE LOAD 1 INCH DEFLECTION
V
SELECTION OF PEAK AND ULTIMATE UPLIFT CAPACITIES FROM TYPICAL FIELD LOAD TEST DATA FOR DRILLED SHAFTS AND DRIVEN PILES (1 INCH=2.54 CM, 1 KIP=4.45 KN)
For this study, field data from 36 uplift loads tests on 27 drilled shafts and 9 driven piles were evaluated where the peak uplift resistance was obtained. The peak uplift resistance was reached at vertical deflections less than one inch in 25 load tests with the remaining seven tests yielding peak resistances at displacements greater than one inch. The ultimate uplift capacity for these seven tests was selected at a vertical deflection of one inch as shown in Figure 3. A method was developed for test foundations where the peak upl ift resistance was not reached during field load tests to estimate the ultimate uplift capacity using normalized curves shown in Figures 4 and 5. The measured uplift load at small deflections was compared to the normalized uplift curves based on the type of foundation and embedded depth to width (D/B) ratio. The ultimate uplift capacity was then estimated for use in this evaluation.
DRILLED
~~ ~ « •.... >zUJ
SHAFTS
AND DRIVEN
120
120
100
~~ --'« •.... > z 100 llJ
«llJ
llJu
llJU 0..«
80
0..«
o~
o~ 0-(/)
D/B
••..•0..
(/)
0-
1.5-2
40
5-8 12-14 16-17
N:J:
:JU
zOz zO «
::;
a:llJ
llJO
:J is
60
::;OZ z0
40
«Z a:llJ
~~
~~
00 ~«
00 --'«
20
•.... 0 11.~
g;o
o
o o
0.4
0.2
VERTICAL
A) DRILLED
FIGURE 4
0.6
0.8
DISPLACEMENT
1.0
1.2
IINCHES)
«
:<:
llJ
:<:llJ
llJ
llJ
> Z 100 «::; llJllJ
o..U
0:5
a..
•••••
llJO
:J is « Z ::;
w
Z 0
o« «•.... ':3 0 •....
C3
11.
--'
o
a:llJ
-
60
Z
Z 0 40 0« •.... « 00 ~ «
40
•....0 ~
20
20
--'
:J a: g;
0..
O~
llJ
o z
IINCHES)
PIERS
-
--'
60
1.2
« •....
> Z 100 « ::; a.. U
z -
1.0
0.8
120 llJ::)
a::
0.6
NORMALIZED UPLIFT LOAD RELATIONSHIP FOR CAST-iN-PLACE CONCRETE DRILLED SHAFTS (1 INCH=2.S4 CM)
120
~ G «::;
0.4
VERTICAL DISPLACEMENT
B) BELLED
•....
80
0.2
o
PIERS
?f..
0:5 a ~ 0
20
•.... 0 11.--' :Ja:
:Ja:
g;o
•....
80
••..•0..
60
llJO
--'
147
:<:::;
:<:::; «llJ
llJ ::)
PILES
~a:
0
g;o o . o
0.2
0.4
0.6
0.8
VERTICAL DISPLACEMENT
A) CONCRETE
FIGURE 5
FIELD
SQUARE
NORMALIZED
UPLIFT
LOAD
1.0 IINCHES)
PILES
1.2
0.2
0.4
0.6
0.8
VERTICAL DISPLACEMENT
B) RAYMOND
STEEL
1.0
1.2
IINCHES)
STEP-TAPERED
PILES
UPLIFT LOAD RELATIONSHIP FOR DRIVEN PILES (1 INCH=2.54 CM)
TEST
RESULTS
Field uplift load tests were performed on 50 drilled piers and 29 belled piers using cast-in-place concrete construction, as well as 10 prestressed concrete and 2 steel step-tapered driven piles. The field load test results are given in Tables 1, 2, and 3 for the drilled piers, belled piers and driven piles, respectively, along with the foundation depth, shaft width, base width, soil types and construction methods.
TABLE
I.
SCt:
FIELD
UPLIFT
LOAD tEST
RESULTS
-
DRILLED
PIERS
.f::.
rORe AYF.RM:F:
PIER OEI'TH
DAn:
!!!lL-l!40 Pile
~
AND LOCATlON
Second
No.
Bould~r-ChJr.o
2-H2UT4
Pi If! No. 4-H5311 Pi I •• No.
~-HMJTJ
Pi 1" No. 6-H69T2 I'i 1•• No. l-Hl?TI Pi
Jr
Ptl
No.
•• No.
No.
1.5
9.9
11.0
1.1
7.J
11.0
1.5
I R.O
1.5
10.0
1.5
9-H97Tl
I J-H20IT/.
Pt1e
No.
I
No.
1.5
q. e
1.5
1'1. ')
I.':i
6.2
1.5
6.7
12.0 Be LL!.L.h 18.0
.2s!.ober I 1941 Chlno-l...,~un. Pl1~ No. I-H149
96.0 (192.0) 7.3 104.0 (200.0) 12.0 IlO.I}) <250.0) 6.7 104.0 ( 260.0) 1,.2 120.0 (187.5 )
9.4
12.0
16-H2J2T2
1.1 I.') I.') 1.5
1.5
11.3 8.0
8.0 R.O
80.0 (Il}.)
6".0 (9B.I) 6".r) ( 10b.1) 120.0 ( 192.0)
N~~4~_~:~~n-Heu Pile- No. 4-H89T4 Pile No. IO-H43T) Pill! No. 11-H4313
ML.-.llli~~~!.!! rn'pakt C()II<:ret •• I'f t£!..!.lLl9S3 I'llI!' No. I
T/1.10•0 10.0 10.0 10.0
10.0
1(0
SCE R~rort
No. 124 10.0
1.66 ).66
'1.0 ( 16.b) '.8.0 (6ij.6)
12.0
No.
10.0
6A
6.0
15.0 12.0
UNIT
FRICTION
WEIGHT
.J.:.L
1::L
I. 99
2.84
I
00
ANGLE
l!s.LL
~~~Al!~
(lX.!p,re~
15.0
1.60
9.36
7ft
15.0
1.60
C).38
I'i If!
No.
7C
I S.O
1.60
9.18
Pi If" No.
1:1
1'•• 6
1.51
Rrrord('d. dtr"('r
IIIhE'lIr
9.57
1.5J
5.53
5.76
5.76
1.36
NE
2.B9
2.15 5,46 3.99 6.26 1.68 2.63 1.39
NE
NE NE
HE
2.J
2.92
1.48
R.74
7.)3
I f,.56
7
5.78
1.IR
3.85
".59
3.68
2.21 Nt:
1.60
2.IS
2.72
3.74
1.29
NE
2.0G
0.1209.
41"
0.\10"
40'
Cf'rnt'oled dR (rOIll 4-11 ft Decompos(Od RfAnite f rOIll 0-7 C"",('nted $find from 7-11 ft
0.125""
10'
Cf'mf'l1t('d
dg
F1llty
Haillt SAndy
dg from 0-4 ft glJt from 4-JR
Holst
dR frnm 0-].';
45'
C".melltt'd fIR frnm 1.5-6 "\lIU~ !ll1ud f lun, 0-1. f t Ccmelll ••d '••wd from 4-10
40'
0.110"
40"
0.125
36'
0.110"
)'):1
:\nd
Loos", U.-Itd
gravel
from
l'illt
and
Clayey TRnnrc
stir top
Sf Ity
9 ..•".1
2.76
42.4
0.130
7.0
(65.1 )
( 1.(0)
0.88
1.21
1.65 1.16 3.12
0.120 0.120 0.120 0.120
31d
51 Jty
I'Inod
with
31d 41d I.Sd
Silty 51 Ity
SAnd
sand
wHh with
Silty
!'lAIrd
w(th
0.100
33b
Loose' sAnrl. Steel C"OI1AtI'"IICt pilE" And
0.126
31 '
CJayey
0.126
31 b
Clayey
0.126
31 '
0.126
31 b
11.8
B)
9.5
0.49 (1.00) 0.111
(1.00) 0.51
(53.5)
( 1.(0)
'!'tltlll,
,)
frnm
SI'T
0.93
1.0
NE
0.54 0.61
0.97
1.61
0.26 0.30
OJ.4
0.11,
0.J6
0.57
0.95
1.00
(51.0) 50.0
1';no!'nthrflt!J
0.41
NR
0.19
NE
NE
0.42 0.~4 0.44 0.54
NE
NE
( 1.00)
'&.5
fn
5.0
NE
0.45 ( 1.00) 0.30 (1.00) 0.52
I S.O
Nltmh,..rtll
4.6)
NE
(l.OO)
(lR.O)
Ob.5)
NE
2.19 2.07 J.12 4.63
NE
NR
0.30
(14 .0)
;HI:" v"llIt.l'I.
(,!'Itlmnt('d d)
from
0.120
0,99 0.94 1. 87 2.78
NR NR
0.28 ( 1.00)
21.3 (2B.
0.10 0.17
1.43
0.)9
0.98
0.'8
N~:
0.52
O.
NE
J
from
rl·IHilt".
O.RS
1.42
l1ormnl17('d (1.0
kJp
rurVf'F1 _ 4.4S
In kn.
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from Aod
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0-7
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ft
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f t
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ft
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9i1od.
gr"'v~I.
31 '
ClI1YC'Y
I
•• 30.48
hAckfil1
,m.
I
Inch"
bllckflll
2.';4
concrete
GHP casing_ tamped l!land CHP casing. flooded 8And CHI' cuinR;. ."nd vlhrllt('d
31 '
'"
uaed tn n'movP\!.
flooded sand <:ttr c8sin.R:. vlhrated "'And
0.122
ClllYPY
casing then
9:1od with 111 around SAnd with
11 b
JI b
oc::'Tj
gravel. .':ravel.
CAst-In-plAce
O. I 22
FI~urf' Font
Clayry
sand.
en
Rr •.•• e1.
0.122
0.122
I
0.76
valu('<;
I. 24
ChyE"Y
It
5z l"'
BAnd.
0-4
backfill around Clnyey sl1nd with bAckf 1\1 Around Clayry S:lnd with hRckfill Around Cll'lyf'Y 811.\(J with
0.87
0.74
!If
RIAvr)
ft
from
Clny('y
bad.f
0.62
0.68
35(:
51 It Y "'and
Aud
ft
tT1
0.39
NE
CI'T
0.86
frol1\
en C/.J
)-10
!'IIInd
1.93
35'
~ ft
t
nnd sIlty 2 feet
S:wd I rom 7-1,
0.110"
en
ft
from
1.1
fI
f
Aaod
e.'m •••,It"'d
!>1Ip:hlly
Z
ft
5-10
I from 0-)
1'1(11
>-
ft
ft
Snnd from 0-5 Sand
~
ft
0.110""
40'
>-3
ft
»"
0.120"
ft
ft
0.1209.
0.120'"
s8nd.
ft
rom 0-4
r
Cemented dR from 3.5-10 L"ose RAnd from 0-1.5
1•• 19
1.22 1.9R
Loolt'
10.75
1.)0
NE
S11Rht1y
45'
N~:
(61.4)
(41.0) 54.0 (\7.S)
I. 91 J.80
3511
0.120"
2.13
34.0
51.0 (\1.5)
3.61 NE
0.1103
l.f,l
21.0 (27 .0) 33.1 (40.1) 29.5
38.0
1.80
0.98 0.19
2B.0
(41.0)
1.47
0.67
6.25
00.0)
0.98
Nt:
6.67
Pi 1 •• No.
frot!>
K
0.130 (1.00) 0.140 (1.00)
1.60
6.2S
b)
37.8 115.4 ) 4'1.8
1.')0
6.2')
•• d.
0.104
19.1
1.60
'·l'Illm,.I
1~
( 1.(0) ( 1.00)
20.9
1.60
NE
72.8
6.67
12.5 (17.0) Ifl.O (21.0) 24.0
O.Olb
(" 1.6)
1.50
10.0
II)
bl.5
61.J (104.01 114.1 (I Rr,.l)
148.0
10.0
Not •• :
REI/!.
FHICTloN
..u:~ HE
(1.00) 0.140 ( 1.00) 0.094 ( 1.00) 0.102
( 96.0)
6.0
6C
Not
(lJo.n
l.fl6
6R
NR -
SKIN
GROUND WATER
0.123 ( 1.00)
0.100
77.4
46.8 97.8 14/, .8
No.
F.nrounten·d.
(I BI.B)
101.0
No.
NF: •. Not
101.1 (197.1) 111.2 (241.2) 101. ) (257.J) 118.3
b.O 6.0
Pill!'
No. ]A.
•
OHI..
~
( 1.00) 0.072 ( 1.00) 0.046 ( 1.00) 0.034 ( 1.00) 0.193 ( 1.00)
(IB9.1)
1.66
Pi}@
rile
(102.7) 9J.1
46.0
131.1) Pile
b8.b
76.0
(II R.B)
(6B.7)
1ii~'
VENT
NET
Q.!l'~
72.5 ( IOb.6)
10.0
1],0
!!~~_~~.!..!2er-f!!..t'!2-!L~. I'JI,. No. 1',-H]24lf< Pt1e
6.3
20.0
~1!.. •.!.U...JJ~~!JE1J
CAPACITY
TOTAL
EF •...ECrtYF.
TIL
PI I •• No. 10-HI09T1.
Pile
UPLIFT
o/s
_~ Q.!l'~
14.8
~-HR9TI
Pil •• No. 11-H179rl
PH:R WIDTH
P;ef>t).
TOTAL
Around CHI' Cllllillg. tI:lnd wHI> tllmped lIanti
UOllnd CHP cuing. C"Rt-ln-pJ,'H~e
"'nnd.
C!I1. I kRf
_
1.7.9
kplt)
('on,n>te.
z o >>-3
5z en
"'<;'i"'~",
rAlH.t: I.
Sl:t: t·IU.I' tll'l.lt'!" LIIAII n,:;1
In-SULI:--
111t!!.:..!....!J!...!:!~~
-
Avt.:RA{;t::
PIt:H DfPTlI
OATt::ANO~!..!!!!
~!..!.l. 1'::1')] Pile
No.9
1'11t:
Nu.
t'11::I<
~IUTII
(feet)
set::
I
Nu.
IUA.
O'ecl
ti
I. 51:1
b.3B
LbO
to.25
12" 10.1 IO.t)
0111
Pile
No.
lOt!
10,0
I.LO
b.25
t'IIt:
No.
IUC
10.0
I.bO
b.2')
t'1Ie
No.
IIA
IU.U
1.)\
b.b)
Pile
No.
lit!
10.U
1.) I
b.b3
Pllt:
No.
11(;
10.0
I. 51
b.b)
t'lle
No.
12
"lie
Nu.
I]
PI It:
No.
llo
(;At'ACITY NET
~cl
~
13.5
b.o]
tJ.1
J.
).bt!
11.5
5.8
( b.5)
(b.5)
9.0 ( 10.0) 22. ') (24.5) b.O
7.0 ( b.O)
20.5 (22.5) 4.2
9.1:1
1.)1:1
1'ilt!
I-H]I'J')
No.
Sc.:ond
SlHl~'~.!.....:._!l<:HfI
9.0
No.
l-HI7T3
5.0
l.l:IJ
Pile
Nu.
)-MIJT2
b. \
1.2)
~!£l!.~Ku-Hlra Site !I]-leg
loma
TII.
C
S.~I:I
~!.(h-SCP(~l!!bCf Site 401
•
I'HU
~v'H,,-t',dll f:I.OO
tJ
SI
te
lo I It)
tI.IJU
4n~-I.I'''''
A
1.1)
tJ
llo.o
"IV
:illa
4.9J
~().)
2./J·) J .110
(b.
IOtI.O
FRICTION ANGU:
u.n ().~lo
u.))
0.11 0.45 0.54
SOli.
JJb
SI Uy
tl.ttnd.
JJb
Silty
",.nd
0.110
3Jb
CHP Silty
0.110
JJb
atound Silty
CHt' undo
0.110
3Jb
aruund Silty
CMt' ca.lng. undo Floo,jed
.ano.l
U.IIO
)Jb
arollnd Silty
IIIIIOOlh IIte!.!1 undo Vlbrnted
cas.ln((. und
U.IIO
JJb
O.IIU
1 .b4
0.))
1.10
J.I)
].1~
2 It. VILrat~d
dl". aand
cdalng. Tamped
und
O.IIU
JJb
.lIIouth Silty
with I>rllird
0.110
))b
IllIO~d Kruut Silty undo below water
0.120a
45"
~~
0.14UOI
55"
dla. b.n .'fflcture"
0.1
50"
dill. failure "one developed. Poorly celll~nt~d .andatooe.
c.alll)( uno.l.
backf11!
arolln.)
holo:. backfill backfill
3Jb
2.52
2.10
In
aand
ateel l·ulr~d ateel IJrllled
0.54
I.]()
cuing lIi1ud.
CllnCrete.
flouded
aruund sll/ooth Silty •••••nd.
O.tO
0.lo9
Caat-ln-phce with
2
AND C()MH~NTS
0.110
O.lob
0.38
CONOITIONS
0.110
O.Slo
O."}U
0.11
~EIGIIT
ili!.L ~
2,02
J.7b
(1.00)
(1)5.5)
O. ]7
(1.00)
90.U (12500)
tJO.1:I (1I5.b)
U.20 (1.00)
bdcktlll bd!.:ktlll
tJ
caalng. .and backlill caaJlIg. 2.0 fet!t Jla. hol ••a 1.6
In t".tto. teet dla.
:;0 hule
anti
to mdlt •• a hole au.)
UIO backf I 11 around IOlliouth Ulled drilling l'uJ tabile.
caMll1g.
uud bell
p rm
tJ
VI
106.0 (110.5) 100.0
1:I'J.4
U. Sb
(9J.8)
( 1.00)
17 .1
0.27 2.50
'/0.1
O. )J Q.lo5
lo.71
ILlob
'J.loJ
2.10
N<
l.)b
, .'1)
t.
1.15 1.57
lo.62
'>.1]
O.7t1 D.b2
1./:I1i
N<
0.59
1.22
2.14 1.45
II
u.71
O.hS
U./:I9
N<
}OOl
.bl
granJ(e.
The
pulled out aandatone.
failure
alollg
O.llU
42c
Sand
U.IO'::l
40J
Silty
••and
o.ln
40b
Ccuwut Silty
slurry lIIaod.
0.120a.
Jbd
"llIrry Silty
to chy
with
of
2 J •.••• :h COllcrele. An 8 feo:t
perllluHer
!:obble.
of
aod
Shear pile
boulder",
~
::J VI
~
Z tJ tJ
:;0 1.9 I.b
13,0 11.9
'J,
10).0 100.0 124.0
(nt!. '>
11/1.u
2l.0
1.15
12.b
Ib.l
120.0 125.0 80.0 45.0 1JO.O 73.0
Ito2.0
0.05 (I.oo)
(2UH.I) 99.0
(1.00) 0.97
10
Ilio.U
O.~O
10
~ .14 O.\J 7.1,0 O.b~
II'L8
U.IO
]).0
"J.9
t •.•••t"'.
c)
1)5.4
o.ll
0.1111"
)]oJ
0.110"
4UJ
II.'III
0./1 O.~<}
1.05
lI.l'.! 0.lJ8
1.21
0.110
,.J
0.93
0.tl6
1.12
U.11O
JJd
10
0.61
0.18
1.2S
0.110
)ld
15
I.OJ
O.tI'i
1.1 tI
0.110
Jld
0. ~ I 0. /19
ICJ
Nl
l.lIO
I.Jl
St'I
art: value
edtlwdted •• ,
0.1)
from
vu.luell CPT
U.l:ltI
I.)t,
I.b8
paro:nthedll>l from
II.).
ICJ
10.00
Ib4.0
NUI~!.t:r •• In
Nl
(i1''!,'I)
1111.0 J)
SUI! lo1"]
tiheu.r
t:Ht::CTIVE
UNIT
:r:
51:1.tI
tI.3
I
1.\
(tJ5.8)
2.0S
dltect
1.5
(bl.O)
17.0
truIU
a.loO
N<
47J9C
b)
1.5
I.b7 O.SS
Site
NI< •• Nut
0.20 ( 1.00)
N<
13.5
tltltlmtttcd,
1.5
I.UU
1.10
a)
1.5
0.09 0.11 0.01 0.08 0.19 0.20
IbO
I. I ~
•. N\,I( t::ncO,Iuntt!rll:'d,
1.5
85.0- tH.9
21.9
Nutes:
l.tH
(1.00)
2to.J
Nt;
0.b1
I.n
(23.b)
47)98
T/I. 1.5
0.")
0.45
( 1.00) 0.52
4]]9A
25.0
O.lb
0.41
IfI.6
Site
•• rl:h.-11~~!~r'lno
I.Ot!
1.5
2U.5
bU.O
O.lll
O.l) 0.26
( 1.00)
])
i:L
0.35
0.10
0.b5
-.1:.L
1.5
19. ) (2J.])
Site
11
~
K
0.12 0.13 0.14
21.0 (25.0)
(2"/.6.2) 2l.1:1 102.0
~y-H Sitt!
O.loO
( 1.00)
9.2 (9.5) 5.2
lolLO
n.
Site "'21 Sit",
Vtlrde TII. 4.21
(1.00) 0.25 ( 1.00)
(J.B)
J
1.3
I.)
0.5)
100.0 /..lJ 10.0 (J]].5)
4.lI
1.5
O.loO
( 1.00) 0.25
1,0
6.19
5.2
U.SO
(1.00)
IH::TA
t'age
TOTAL
SKIN
t'l
(f.'eet)
J.2
11.0 (11.2 )
T/!:.._~~ll~~~. 1.10 5.29
"lIe
~ATl::R
( 1.00)
0.])
(2505) ~~~~~!
C!WUNU
HEFL.
S.O
( b.O) 5/:1
VI::N.T.
~
(I J..)
(1500) 1.t!
(1.0)
1.'1
IU.U
lJl-'LlfT TOTAL
trulD
result'"'
lIoru'dlll-ed
(I.O
k1"
curve.:! •. 4.loS
0.12ua
1.41
In kn,
.·Ig:urt:
4.
I t'out
•. 30.41:1
)4J
em,
I
Jnch"
with
cobblea.
tu ralnimhe U"ed cement
"Inlmh:e. & _t lty
cavin8. .an,l.
U.ed cavJng.
tal.I.,
sand
Silty drilling
ulld
Stlty drilling
und
SI Ity
.land
2.54
Cal,
m
Z Uud
drlllln8 •••"d bel()w wal.r tabl •• Silty aalloJ ••,ui clay. Uaad ,IIIIIII'K •••",j t.•• luw wat.r Sa 11.1 11110.1 elllY .and. Ua.d dr111111& .ud b~luw water tabla. Sf lty und and chy. Uaed drlll1n¥ SlJty drlllJng
=2
'"0
P
r;j
IImoJ b.,low water tabh. /lild clay. Uaed .ud below water table. and clay. Ultld lIIud below Watar tabla.
Ind
dlY.
IIl1ld below
I kat
and
_
Uud water table.
clay.
41.9
kpa)
~
\D
150
TRANSMISSION
LINE TOWERS
FOUNDATIONS
..: ~
to;
;>
>
or
It'
0:.
a,'
;>
::>
ce
0:
co.::.t!>t!>t .,r..;;;.;;;.;;;
3-;3-;
::~
-;:;-;:;":"C
~
>->->.>.
~o co
~u c
~
..:
= •••. _~C>"
0'0-"''''' c::::::c
=:;x:,~ •...••.... _ ~:t;::~~g:::,~ _.cc..::...:-X ..:.c.:.~o= ~cooo-=oc
........................
~O""'OOONC 0-::;'-0-0............................
~~~~~~~~ ~~g~gg~g~g;~~ ............. =C-:::--N-"'-~=---. N •.•.••..••.•. ~OOc ~:i~:~:i~~ ......
coc
~~~~~~;:~ c-oco-oo
•.•...• .:cO•...• NON:::
•.•. <:OO.D •.••• :::=:::. 00":''':'0'':'0''':
~~:~ =~=~
.
~~;::,::;:;:;
.......................
.............
~"':=:r.:":~~=:oo~='2='o=cc=c2
.
co:::o..cc~:: . . . . . . . coocoooc
•.... ..:"., •••• ::c.c-~c•.... o.c •....• =~~ ~:;:~~~~£o~~ :::::;.~ •.•."'..., ••.••.•.•••.• co.:: ==co o,o,~= 2giii;:~ - --...•."....,. - ••.••.....• -- - ...,
•..•
•...
~ z
.
~c ;:
TI::; pcC"H'
............•••.•.•.........••••..
lli!.!.L~1:l l!.!f.!l~
4.1 1'11::k. Gk,UUHO ~ATt::K ISHI. fRICTION SKIN UHI tI0.61 vt:k.T. CA1'ACl1'Y Silt UNIT TOTAL \.IIUTII K ANGLE 4.3 74.0 2.0 40.0 05.0 3.6 3.0 3.5 3.1 1.0 2.71 2.0 0.13 0.4) 0.69 0.54 2.49 1.16 0.99 3.40 0.66 0.&0 Ion 1.00 EHt:CTlH t'k,l N< 46(NE 1.51 Dt::FL. 2.25 N< J0.61 N!:::T 1.49 3b1.11 6.0 9.4 7.4 1.4 6.1 l.4 ISI-.:TA CTION TOTAl. ~t:ICIiT 0.100 1.60 0.55 1.41 1.20 SUt 8.25 1.1 SOIL 0.126 0.114 CONIHT(ONS AND COI1Ht:NTS TABU: 2. 31.0 56.0 1.0 ~loTIl 44.0 ••0.02 (0.52 0.0 1.34 1.00) (36.4) NE (12), 1.12 1.00) 161.4) 30b 33b U/IS 0.114 2) 14.0 6.5 0.115 Silty und with cobbl •• 1.!.!!£h1 sc~ FIHD UPLIft LOAD n:ST RESULTS - BULEO PII::H.S (Cont.) (Fel!!t) ....i=L (t'cet) -1!!lL 1:L Silt Sihy und with .-f!£!..L ~ cobblu )lb 31 b )2b Avt:RAGI:::
)J<
o::0 Ht: •• Not t:ncountored,
NR •• Hot Rl:!cord~d.
Note,.;
b) troll
NUlllbt:ul in parenthl:llill
lire e,tillilHed
YMluea frolll normalized
curveM
in Fll1ure
F r
4.
tT1 a) t:llltim.l.rad,
KTucker02:
direct
aht::ilT talllta.
c)
Crom C1'T raauit
••
(1.0
kip"
4.45
kN. I Foot
•• 30.48
cm,
I Inch"
2.54
em, I kat
•. 47.9
kh)
o (/J
nptl3
::r:
illlli ~ ~ A
TAbU::
Predrll1ed toSand ~231.1 feet depth. GttulJNO p(I,t: fRICTtoN EfFJ::CTIvt: WIUTU SKIN UNIT TOTAL K 0.66 TOTAI. WAHN 210.0 110.0 120.0 212.0 101.0 150.0 o. 1.05 0.16 0.69 O.b4 O. 0.15 201.7 0.90 0.56 0.71 0.55 1.28 0.84 1)).71.03 0.63 203.6 0.~4 O.JI) 0.65 O.bb t'k 0.80 0.02 0.12 ANCLE 15 14 46.9 49.7 46.6 30.4 32.6 31.7 WI:::ICUT 34.0 34. 0.105 7~ N£T 0.63 0.74 J7 0.)1 O 00.55 DEn. 0.31 0.52 0.99 ICtI 00.9J 0.10 0.63 .~3 .9~ .54 .41 ))" .80 )tlt:TA ON upLin 0.64 0.66 SOIL Vt:kT. 56.0 3').0 CAPACITY 26.7 54.0 and CONUITIONS lIih. .tlt. 0.1101:1 ANI>COHHt:NTS 215.0 90.0 0.45 63.9 42.4 34.3 0.66 D.H 0.9b 40.0 136.0 141.0 70.0 IloJ.9 (217 66.4 (167.5) 164.9 14 11).8 (246.1 208.1 (221.6) (230.3) 0.66 10 91.0 26.7 0/. 0.122 0.110· 91:2. 0.10 0.~2 0.40 0.42 (0.65 ))b .59 .J) 1.(0) ) .Predrllled (0.69) 0.96 1.2040.0 49.) 34 51.0 .0totu.thy 0.110 0.122 (224.0) 139.6 1.11 1.17 1.00) IIIq. Mq. IiIq. Predrill1nl Silty Pr.dr1l1ed NO No Predrilled predrUl1na. predrlll1ng. aand lIand )8 Pudrtlled Pr.drilhd )4 andfut to ailt. .Ilt. 30 2) depth. dal'th. feetto depth. 39 40 fut (14b.lo) 1.1 J3)b .•• ilq. predrillinjl. 0.110'" I'udrllhd 42 het hilt 1.00 ('·ect) ....i=L .q. 8q. ----1!!.1L i!::!.ill 40< S.nd and ) tut .17 aand. sq. Sand ••. ndq IIllly lIand. SlJty and 8tlt)' IIlit. .and. unal Alld 11lIt. Silty Stlty und •• nd and ,ilt. .Ilt. at predrflHna:. It. fu.!..L 31b 33b r~::~
I.)) ~
1ill!l.
J.
SCt: fino
Unin
LOADn:sl'
>
kt:SUI.TS - URlvt:N PIU:S
::J
AVt:RAGt: U.'.ItI
dapth. depth. d.pth.
(/J
40< )1b ))b
1'1190
~
o o::0 =2 tT1
Z 'l::J
F tT1
(/J
NE •• Nut Encountered,
Hit • Not Recorded.
Not~y:
b)
a) utimated,
frOID
direcl
Numbers
aha ••r le&tll,
tn p.lHenthealll c) froll! CPT (aIlU!tV.
esthlated (1.0
valuell kip
•• 4.~5
from normalized kH,
1 foot
curVeM in Figure - 30.48
cm, I tnch
4. • 2.~4
CII,
I kilt •• ~7.9
kl'a)
VI
152
TRANSMISSION LINE TOWERS FOUNDATIONS
UPLIFT BEHAVIOROF FOUNDATIONS The uplift behavior of drilled shafts and driven piles depends upon the foundation geometry, as shown in Figures 4 and 5. Drilled piers exhibit a cylindrical shear failure surface along the soil-pier interface that is mobilized at vertical deflections of 0.25 to 0.8 inches (0.6 to 2.0 cm). The longer piers reached peak uplift resistances at greater displacements than for short piers, as shown in Figure 4a. The belled piers yield a complex failure surface depending upon the base configuration and in-situ stresses. Short piers with D/B ratios less than 3, in normally consolidated deposits, mobilize an enlarged cylindrical shear surface with the peak uplift resistance at vertical deflections from 0.4 to 0.7 inches (1.0 to 1.8 cm). For belled piers having D/B values from 3 to 5, an inverted cone failure surface develops at larger displacements from 0.8 to 1 inch (2.0 to 2.5 cm) or more. The longer piers with depths greater than 6 times the shaft width yield a general cylindrical shear surface which occurs at vertical deflections of 0.0 to 1.0 inches (1.5 to 2.5 cm). Generalized failure surfaces for belled piers at various D/B ratios are shown in Figure oa.
r
100 BELLED PIERS *0 " ISCE FIELD TESTS ON DRILLED PIERS FIELD TESTS ON CONE BREAKOUT I! b I SCE *0 0.3 FIELD I TESTS q I TEHTATIVE STONE AND lIMIT~ WEBSTER I OF 1.0 I0.4 19 * 001.0 0.50.5 at ",0 ROCK I *o! 1·~.~0.5 1.06 ,61.0 I ' *0 * I MODELi TESTSI I /" VALUE OF ZlD DlJP ! 0.2 I *0 I ~1·:1
*
-H-
Q
I\//\II \
I
D/B
r
>6
Il-Z -'" Q52 « a: CI) ()
I\
--
IQ\ /\
1\J
i= 1L1L 0()z0~,I -.0 c:
o
o
~8
«z
10
D/B=3-6
0.1
12 EMBEDDED A)
GENERAL
FIGURE
6
FAILURE MODES FOR BELLED
COMPOSITE
FAILURE
SURFACES
PIERS
FOR DRILLED
B) CONE BREAKOUT
SHAFTS
DEPTHI AVERAGE CHART
16
WIDTH-D/B
FOR DRILLED SHAFTS
IN UPLIFT
In granular soils with high in-situ stresses, the cone breakout was noted for drilled shafts with D/B ratios of 6 or less, as shown in Figure ob. Drilled piers in these overconsolidated deposits may deve lop sha 11 ow cone breakout patterns in the upper porti on of the foundati on. The belled pi ers produced an inverted cone surface from the enlarged base up to the surface with radial cracks observed at higher displacements.
DRILLED SHAFrS AND DRIVEN PILES Driven piles displacements curves in perimeter deflections D/B ratio. FACTORS
with embedded to obtain the
depths up to 50 feet peak uplift resistance.
153
required larger The normal1zed
Figure 5 show the soil-pile failure surface along the of the driven piles was fully mobilized at vertical of 0.4 to 1 inch (1.0 to 2.5 cm) or more based on the
INFLUENCING
UPLIFT
CAPACITY
The uplift capacity of drilled shafts and driven piles in granular materials is influenced by the shear strength and stress history of in-situ materials, foundation geometry, construction methods and other parameters described in detail by Kulhawy and others (3, 5, 7, 8). The expanded general equation for side resistance is expressed as:
(3 )
with wh ere
K
original vertical friction
fs
=
(ovl)(Ks)(tan
( 4)
&')
operative coefficient of horizontal soil stress, Ko in-situ coefficient of horizontal soil stress, 0v' effective stress and &' = effective stress angle of for soil-shaft interface. =
The skin friction factor, Bs, is the single parameter which incorporates these factors with the effective overburden pressure by the following relationship:
( 5) Shear strength - The shear strength of the granular soils were obtained from drained direct shear tests on selected samples as well as correlations with field SPT and CPT results (3, 11). The effective stress friction angle, <1>', was selected for each site and used in evaluating the coefficient of horizontal soil stress, Ks,
using
the Ks
following =
Bs/tan
relationship:
&'
=
fs/(ov')(tan
&')
(6 )
A detailed study (4) of soil-concrete interfaces has shown that with normal cast-in-place concrete placed yielding a rough interface, 0' ::: <1>'. The use of steel casing reduces the roughness along the soil-shaft interface with the following results from Downs and Chieruzzi (1) are given in Table 4.
ed
TRANSMISSION LINE TOWERS FOUNDATIONS
154
TABLE
4
EFFEC1
~
OF STEEL
CASING
ON UPLIFT
CAPACITY
Method
lli.ill =
30.48
SHAFTS
Ground
0.38 NE Water 6' 0.83 1.10 0.35 0.8& 0.&--L:.L O. 0.90 &4 1 pIers 1.5 SteelCasIng CMP Belled DrIlled pIer 1.5
(1 foot
OF DRILLED
of ConstructIon
cm) Flooded
sand
VIbrated 1amped
backfIll
sand sand
Concrete
backfIll
backfIll
grout
around
placed
steel
around
usIng
casIng.
steel
casIng.
tapered
around
mandrels
steel
casIng.
Flooded sand backfIll around steel above belled portIon of foundatIon.
casIng
From these tests, using steel casing for dri lled shafts above the water table, 6'/¢' values range from 0.6 with nominal compaction of backfill materials up to 0.9 when a high level of compaction was performed. The 6'/¢' values for similar shafts placed below the water table range from 0.35 with flooded and lightly vibrated granular backfill materials up to 0.85 for soils compacted with driven mandrels. Also, the use of cement grout around the steel casing yielded a to results from
6'/¢' ratio Kulhawy and
granular soils. For assumed for drilled construction. stress increase
greater Peterson
than (4)
1.0, which for various
is similar grouts in
this study, a 6'/¢' ratio of shafts utilizing cast-in-place
unity was concrete
History The original in-situ or decrease due to method of
soil stress, Ko, construction, changes
may in
overburden pressure, cementation and time. The in-situ stress history of the granular soil deposits may be estimated using results from pressuremeter tests or empirical correlations with field and laboratory test indices. Studies by Kulhawy, et al (3,5), have shown that the K/Ko ratios 1 when normal cast-in-place
for dri lled concrete was
shafts used.
vary
between
Foundati on Geometry was evaluated for
- The embedded drilled piers,
Drilled surface
driven piles yielded along a cylindrical deflections from 0.25 to 1.0 inches (0.6
piers and at vertical
depth to shaft wi dth rati belled piers and driven
2/3
and
0,
D/B, piles.
shear to 2.5
cm) or more. For belled piers with D/B values greater than 6, the failure mechanism may be approximated by a cylindrical shear surface using the mean shaft width from the following relationship: Bm = Bshaft with Bbell
+ 1/3(Bbell
Bm = mean width of belled = width at base of pier.
( 7)
- /Bshaft) pier,
Bshaft
shaft
width
and
EFFECTIVE
DRILLED SHAFfS AND DRIVEN PILES
155
Construction Methods - Drilled shafts placed below the groundwater table were constructed using drilling fluids to minimize caving and sloughing of the granular soils. A thin film or thick cake of slurry bui lds up along the soi l-shaft interface whi ch reduces the uplift ratios
capacity of the foundation. Previous of 2/3 for this type of construction.
casing shafts.
and
EVALUATION
groundwater
OF SIDE
The load test and laboratory relationship:
Qs and
reduced
were data
uplift
capacity
of
drilled
evaluated with field exploration records using the simplif1ed side resistance
oJ~As)(av')(Ks)(tan Qs
the
KIKo steel
RESISTANCE
results test
=
also
studies produced The 1nfluence of
= I1/2
B D (av')
(8 )
~')dz (Ks)
(tan
(9)
~')
For concrete for dr11led shafts with KIKo = 1 and o'/~' = 1. square piles, the constant I1 should be replaced by 4.0 to evaluate for f1eld load test results. The mean width from eq. 7 was uti11zed belled piers. Average by the surface
Skin Friction, fs - The side resistance, Qs, was embedded surface area, As, assuming a cylindrical for drilled shafts and driven piles to obtain the
skin friction value, fs. higher shear strengths, as 20 , 64 45 35 8 1 a55 1.0 en 0.4 0.8 Zw 0.6 en 5~ ao2: c••.. u2:o II'IE" MOVE I25 Q u: '" ::; I 10~ ¢'•(DEGREES) cwu.'D •••• BELOW FRICTION ANGLE0.2
a: CD u.. w
0«z
The average skin friction shown in Figure 7a. ,
divided shear average
1ncreases
with
20
FOUNDATION TYPE 6 DRIVEN BEUED P'llE "'fR WATE" TMl.E WATER TMLE
Z u..
::; en I
1 0.2
6 4 10 84 62 2 20I 810 40 u: w a5Q : 0.6 CD 0.8 DRILLE 2 I~en w 0.4 . 10 LOWER I $!
«z
EMBEDDED
FIGURE
7
VARIATION FOUNDATION
OF AVERAGE GEOMETRY
SKIN FRICTION,
DEPTHI AVERAGE
fs,
WITH
SHEAR
WIDTH
= D/B
STRENGTH
AND
TRANSMISSION LINE TOWERS FOUNDATIONS
156
Meyerhof
(6)
gave
average
skin
friction
values
for
driven
piles
in
granular soils which overestimate the fs values from SCE tests on concrete and steel step-tapered driven piles. Most of the SCE driven test piles utilized predrilling operations to minimize driving stresses in the piles. Predrilled holes reduced the in-situ soil stresses and average skin friction along the soil-pile interface. The average skin friction decreased materials as shown in Figure 7b.
for larger The cone
D/B ratios in similar breakout surface and
enlarged width for drilled shafts in dense soils values for shorter piers. Also, cemented sands, materials yield average skin friction values from (190 to 575 kPa). Skin Friction Factor, Bs which incorporates the may be easily computed
-
The use of a skin friction factor, Bs, in-situ soil stresses and a'/¢' factor from eq. 5 once the vertical effective
stresses are obtained. A limiting value of Bs was incorporated for load test results in rock and ma t e ria 1 s, ass h own in Fig u re 8 a . increases with higher shear strengths becomes Figure
with
z z0
10 in rock, as shown in noted for drilled shafts
(5).
,
III,,24 I {3, 45 L-0.4 I6 -55 Kp TAN ¢/ WATE" 20 35 I , 0.6 0.8 e!). u.. n TABLE• fIIBOUND f,BELLED"fR - [AVEN :x:: c:6.c: D ,r/) (/) 10P"M.E 1I II ~ ! «25 UPPER « 81 tU = . 1.0 Q 0.2 ~!D T I 0 =-~:DT~~ -..0 ¢' !DEGREES) FECTIVE FRICTION ANGLE-
z
(tan ¢') granular
= (Kp) cemented
ski n f r ic t ion fact 0 r , B s ' decreases as the D/B ratio
The and
values may exceed larger. The Bs 8b, and cone breakout surfaces were
Bs>l
increase the fs gravels and rock 4 to over 12 ksf
"
DAIlLED
"'EIII
20
BElOW
AaOVE -..0
Z
10
~
8
~
6
?-
4
~
2
I
of-c:
1
o
u..
z 01
Q
o
1.0
a: 0.8
z
i I
u..
LOWER BOUND FOR DRILLED SHAFTS
0.6
~ (/) 0.4
I LIMITS FOR DRIVEN PILES
I
0.2
1
2
4
6
I
I I
I
8 10
20
EMBEDDED DEPTHI AVERAGE WIDTH
FIGURE 8
=
VARIATION OF SKIN FRICTION FACTOR, 115, WITH SHEAR STRENGTH AND FOUNDATION GEOMETRY
D/B
DRILLED SHAFTS AND DRIVEN PILES
157
Coeff1c1ent of Hor1zonta1 S011 Stress, Ks - Once the vert1ca1 effect1ve stress and effect1ve fr1ct10n angle of 1n-situ s011s are selected, the coeff1c1ent of hor1zonta1 s011 stress, Ks, was obta 1ned from eq. 6. The Ks va 1ues ranged from 1.2 to 4. a for dr111ed shafts 1n granular 50115 and were as h1gh as 10 at the rock sites shown in F1gure 9a. A 11mit1ng value of Ks equal to Kp with Kp = 1+s1n -i6 35 I 42 ~:.:: i-wO 25 , OU) .,... '"DRIVEN MEYERHOF ~U) 0U) (1976) LL0" w~ () 00 FOR PILES LL, z~ LL..J ..J ~=:: i-U) 0.4, / I 0.2 I
20
I ! lot.
10
10 ..J
«i--
z>
6
N-.c
4
o~
CEMENTED SAND. GRAVEL & ROCK
8
o
i3 i~(1 I _.
0: 0
Ou
UPPER BOUND FOR DRILLED
:.::2 OU) LL,
I SHAFTSI
i-U)
Zw ~ (:
FOUNDATION
o •
Q
a
TYP"E
DAH.LED "'fA WATER TMlE
ABOVE
DRILLED "'EA BELOW WATER TASLE BELLED flfER DRIvEN
45 EFFECTIVE
FRICTION
¢'
(DEGREES)
U)
] j
I'"
~
I
te
1.0
0.8
tt w :::! 0 0U)
06
()
0.4
~
I
.
P'lLE
55 ANGLE-
FIGURE 9 VARIATION OF COEFFICIENT
0.2.
40
1 EMBEDDED
DEPTHI AVERAGE
WIDTH
= D/B
OF HORIZONTAL SOIL STRESS, Ks, WITH
SHEAR STRENGTH AND FOUNDATION GEOMETRY
CONCLUSIONS Southern Ca1iforn1a Edison has performed more than 100 field uplift load tests on drilled shafts and driven p11es over the past 50 years along transmission line routes and at various facilities. Results from 91 field load tests were evaluated to provide corre1at10ns with field exp10rat10n records and laboratory test indices for comput1ng the ultimate uplift capac1ty of drilled shaft and dr1ven pile foundations.
TRANSMISSION LINE TOWERS FOUNDATIONS
158
A deflection criteria based on 1.0 inch (2.54 cm) vertical displacement of the foundation was utilized to obtain the ultimate uplift capacity from field load test data. Normalized curves were produced from 36 uplift load tests in which the peak uplift resistance was reached. The estimated peak or ultimate uplift capacities were computed for the remaining 55 load tests using these normalized curves with the type and D/B ratio of each foundation. The results are given in Tables 1, 2 and 3 for drilled piers, belled piers and driven piles respectively. The shear strength parameters fs, Bs and Ks were obtained from equations 3 through 6 utilizing the average shaft width for drilled piers and driven piles, and the mean shaft width from eq. 7 for belled piers. The average skin friction, fs' skin friction factor, Bs, and coefficient of horizontal soil stress, Ks, were compared to the effective stress friction angle, 4>', as well as embedded depth to width ratio, D/B, with relationships shown in Figures 7, 8 and 9, respectively. From the SCE field test results, each of the shear strength parameters increased at higher values of 4>' and decreased as the relative depth of the shaft became larger. Drilled shafts constructed below the water table with drilling mud gave lower bound values of fs' of steel casing based on the adjacent to the
Bs and Ks. The presence of groundwater and use may reduce the uplift capacity from 10 to 50 percent compactive effort in granular backfill materials shaft.
SCE field load test results on driven piles were compared to relationships from Meyerhof (6) in Figs. 7, 8 and 9 for driven displacement piles. The corresponding parameters from SCE tests are quite low, due to predrilling of small holes prior to pile driving operations. Methods to predict the uplift capacity of driven piles from CPT records provide good correlations for the SCE test results in saturated materials with low relative densities (9). For
drilled
shafts
in
cemented
sand,
gravel
and
rock
materials,
the
use of a limiting value for Ks equal to Kp was adopted for higher shear strength values of 4>' ~45 degrees. Previous tests on steel step-tapered driven piles in sands (2) yielded similar results where the in-situ horizontal soil stress approached the passive earth pressure coefficient, Kp. Also, drilled shafts in soils with high in-situ stresses (Bs>l) produced a cone surface for D/B ratios of 6 or less.
granular breakout
ACKNOWLEDGEMENTS The author wishes to acknowledge the support of SCE engineering and construction personnel in conducting the field load tests. Mr. Robert Burks, Manager of Civil/Hydro Engineering, and Mr. Shahen Askari gave valuable input and support in preparing this paper. Also, Professors Fred Kulhawy of Cornell and Jean-Louis Briaud of Texas A&M provided insights and reference data for use in evaluating the field test results.
DRILLED
SHAFTS
AND DRIVEN
159
PILES
REFERENCES
1.
Downs, D. 1. and Foundations,lI Journal Paper 4750, April, 1966.
R., IITransmission Division, ASCE,
No.
Tower 92,
2.
Ireland, of the Foundation
3.
Kulhawy, F. H. and Peterson, Interfacesll, Proceedings of Soil Mechanics and Foundation 1979.
4.
Kulhawy, F. H., Trautmann, C. H., Beech, J. T. 0., McGuire, W., Wood, W. A. and Capono, C., Line Structure Foundations for Uplift-Compression Report EL-2870, Electric Power Research Institute, California, February, 1983.
5.
Kulhawy, Proceedings Mechanics California,
6.
Meyerhof, Foundationsll, ASCE, GT3,
7.
Reese, L. C., Touma, F. T., and O'Neill, M. W., IIBehavior Drilled Piers Under Axial Loadingll, Journal of the Geotechnical Engineering Division, ASCE, Vol. 102, No. GTS, May 1976.
8.
Stas, C. V. and Kulhawy, F. H. IICritical Evaluation Methods for Foundations Under Axial Uplift and Loading, II Report EL-3771, Electric Power Research Palo Alto, California, November, 1984.
9.
Tucker, Datall,
Special
H. 4th
Chieruzzi, of Power
F.
0., IIPulling Tests on Piles in Sand,1I Proceedings International Conference on Soil Mechanics and Engineering, Vol. 2, London, England, 1957. M. S., IIBehavior of Sand-Concrete the 6th Pan American Conference on Engineering, Vol. 2, Lima, Peru,
H., IIDrained Uplift Capacity of the 11th International and Foundation Engineering, August, 1985. G.
G., IIBearing Journal of the March, 1976.
F., O'Rourke, IITransmission Loading,1I Palo Alto,
of Drilled Conference San
Capacity and Settlement Geotechnical Engineering
Shafts,1I on Soil Francisco,
of Pile Division,
of
of Design Compression Institute,
K. 0., IIUplift Capacity of Pile Foundations Using CPT Proceedings of the In-Situ '86 Conference, Geotechnical Publication No.6, Blacksburg, Virginia, June, 1986.
10.
Ves i c, A. S., Site, II Journal ASCE, Vol. 96, 561-584.
IITests on Instrumented Pil es, Ogeechee Ri ver of the Soil Mechanics and Foundations Division, No. SM2, Proc. Paper 7170, March, 1970, pp.
ll.
Vi llet, W., and Mitchell, J. M., IICone Resistance, Relative Density and Friction Anglell, Proceedings of the ASCE Session on Cone Penetration Testing and Experience, St. Louis, Missouri, October, 1981.
Foundation
Design for Directly Embedded
Single Poles
by Richard A. Bragg1 Anthony M. DiGioia, Jr., Fellow Vito J. Longo3
2 ASCE
Abstract An improved model has been developed for foundation analysis/design of directly embedded, single-pole electric transmission structures subject to high overturning moments. The model uses a multi-spring, nonlinear subgrade modulus approach to predict the load-deflection response and ultimate capacity of direct embedment foundations placed in multi-layered subsurface conditions, and with uniform or multilayered annulus backfill. To verify the predictive capabilities of the model, ten full-scale lateral load tests were conducted on directly embedded transmission poles. The development of the subgrade modulus and bearing capacity expressions are described. Comparison of the field load tests, and model predictions of the ultimate overturning moment capacity and load-deflection behavior are presented. Introduction Directly embedded single wood poles have long been used by the electric utility industry in the construction of distribution and transmission lines. However, wide spread use of directly embedded wood, concrete or steel single poles for the construction of more heavily loaded transmission lines has, in general, been limited. This is mainly due to a lack of basic knowledge concerning the performance of the directly embedded poles subjected to a high overturning moment at the ground line and due to the lack of a design methodology for computing the ultimate capacity and load-deflection behavior of the embedded portion of the transmission pole which has been verified with well-documented load test data. This paper presents an analytical model suitable for the analysis and design of direct embedded pole foundations subject to lateral loads (combination of moment and shear). The model was developed by modifying the four-spring nonlinear subgrade modulus model for drilled shaft foundations developed for the Electric Power Research Institute
1project Engineer, GAl Monroeville, PA 15146. 2president, 15146.
GAl
Consultants,
Consultants,
Inc.,
570
Inc. ,
Beatty
570
Road,
Beatty
Road,
Monroeville,
PA
3project Manager, Electric Systems Division, Electric Power Research Institute, 3412 Hillview Avenue, P.O. Box 10412, Palo Alto, CA 94303.
160
161
DIRECTLY EMBEDDED SINGLE POLES
(EPRI) under Project RP-1280-1 (1) and described by DiGioia, Davidson, and Donovan (2). A field testing program, consisting of 10 full-scale foundation load tests in soil, was conducted to test the predictive capabilities of the modified model. The development of the direct embedment foundation model and comparisons of model predictions with the observed field load test results are presented. Review
of the Four-Spring
Drilled Pier Model
Direct embedment foundations may be described as a cylindrical shaft type foundation constructed by augering a hole in the ground, inserting the transmission pole, and backfilling the annulus between the surface of the pole and the in-place soil. Due to the similarity in geometry, loading conditions, and the mode of resisting applied loads to drilled shaft foundations used to support single pole type transmission structures, the four-spring nonlinear subgrade modulus drilled shaft model developed for EPRI Project RP-1280-1 (1) was selected as a starting point for the development of a direct embedment foundation design/analysis model. Referring to Figure 1, the four-spring subgrade modulus model characterizes the soil-foundation interaction through the use of four discrete sets of springs. Lateral translational springs are used to characterize the lateral force-d~placement response of the soil. Vertical side shear springs are used to characterize the vertical shear stress-vertical displacement response at the perimeter of the pier. A base translational spring is used to characterize the horizontal shearing force-base displacement response, and a base moment spring is used to characterize the base normal force-rotation response. Figure 2 shows schematic representations of the various springs and gives expressions for the corresponding subgrade moduli. Since, the load-deflection relationship for laterally loaded drilled shafts 'is highly nonlinear, the relationship between lateral pressure and deflection was modeled using a variant of the so-called p-y curves developed by Reese (3) and his coworkers at the University of Texas. Referring to Figure 2a, the resultant nonlinear p-y expression for the lateral translational spring is (1): 2khy p = 0.6 Pult
(1)
( Pul~ )0.5
wher: Pult is the' -ultimate lateral bearing pressure and kh is the lateral subgrade modulus. The other three springs of the four-spring as shown in model were modeled as elastic-perfectly plastic Figures 2b, 2c, and 2d. The ultimate lateral capacity for the four-spring model was determined using a methodology similar to that proposed by Ivey (4), but incorporating the ultimate lateral bearing capacity theory of Hansen (5) to determine-the ultimate lateral pressure, Pult' in the above p-y expression. The ultimate vertical side shear moment is derived from the vector resultant of vertical and horizontal shearing stresses corresponding to the fully mobilized shear strength at the
TRANSMISSION
162
LINE TOWERS
FOUNDATIONS
'LA TERAL
y- TRANSLA SPRING
.. ' TIONA!.. (typ)
-VERTICAL SIDE SHEAR MOMENT . SPRING (typ)
.. CENTER OF ROT A TION ......................................... ........................................ ........................................ ......................................... . .... -
s·"A:t; E" M' 6tie Nt"·S PR i"N'" ..........................................
..........................................
Hun~i~ kb -:TRANSLA BASE SHEAR TIONAl SPRING
FIGURE l.--Four Spring Subgrade Modulus Model
P
Pu1t 2
1--1 (. Jktt=\fJ
\
-04 (O/B)
I
re= 0.55 E B
.
y
(A)
LATERAL
SPRINGS
(B,) VERTICAL
SIDE SHEAR SPRING
I'rk8b = 0.24 U (C)
BASE SHEAR SPRING
FIGURE 2.--Schematic
E B(D/B) 0.4
(D) BASE MOMENT SPRING
Representation
of Springs
DIRECTLY
EMBEDDED
SINGLE POLES
163
pier-soil interface. The ultimate shearing force and moment at the base of the shaft were determined from an equation of vertical equilibrium combined with assumptions concerning the percentage of the base in contact with the subgrade and the distribution of the base normal stresses (1). The model described above was incorporated into a computer program PADLL (~ier Analysis and Design for Lateral Loads) (1) which has geotechnical design and analysis capabilities for drilled shafts subjected to high overturning moments and lateral loads and embedded in multi-layered soil profiles. Proposed
Model for Direct Embedment
Foundations
The major difference between the geometry of direct embedment foundations and drilled shaft foundations is the presence of the backfilled annulus surrounding the perimeter of the direct embedded structure. The influence of this material on the stiffness and ultimate capacity of the lateral translational spring and the vertical side shear moment spring must be considered when the strength and deformation properties of the backfill differ from those of the surrounding soil. Consequently, the four-spring drilled pier model was modified for direct embedment foundations by adding two addi tional spring sets. A lateral translational spring and a vertical side shear moment spring modeling the load-deflection characteristics of the annulus backfill were added in series to the previously existing lateral translational spring and vertical side shear moment springs of the drilled shaft model. The relative contributions of the four springs to the load resistance of 14 prototype drilled shafts tested during EPRI Project 1280-1 were determined (1). Based on the results of this study, the base shear and base moment springs were determined to provide only a relatively_small contribution to the overall stiffness/ ultimate capacity of the drilled shafts. Therefore, these springs have, for the present, not been included in the direct embedment foundation model. Figure 3 provides a schematic representation of the revised four-spring model for direct embedment foundations.
the
Subgrade Moduli.--In the case of the lateral translational spring, nonlinear pressure-deflection relationship given by Equation (1)
was maintained. However, the subgrade modulus, kh, required revision to account for the presence of an annulus material having a different modulus of elasticity (Ea) from that of the in-place soil (Es)' Figure 4 presents an illustration of a direct embedment foundation in cross-section. When E a equals E s the combined stiffness of the annulus lateral spring and the in-place soil lateral spring should approach the stiffness of the corresponding lateral spring for a drilled shaft of diameter Bo' When E a is much greater than E s , the combined lateral spring stiffness should approach the lateral spring stiffness for a drilled shaft having a diameter of B.
164
TRANSMISSION
LINE TOWERS
Q
FOUNDATIONS
/1;\M , ANNULUS
LATERAL
SPRING
IN-PLACE NATURAL SOIL LATERAL SPRING
ANNULUS VERTICAL FORCE SPRING
RIGID LINK
IN-PLACE NATURAL SOIL VERTICAL FORCE SPRING BACKFILLED
ANNULU
if ~f~
BASESHEAR MOMENT SPRING BASE FORCE SPRING
FIGURE 3.--Direct Embedment
Foundation
~ DIRECT POLE
Model
EMBEDDED
NA TlVE SOIL
BACKFILLED
FIGURE 4.--Cross-~~ctlon
of Direct Embedment
ANNULUS
Foundation
Using these two limiting conditions and the concept of combining the annulus and in-place soil springs in series, yielded the following expression
for the annulus spring stiffness
a
Ea
1 -
(D/B
0)-S
(BIB o )-S
(Kha):
(2)
DIRECTLY and
the
following
EMBEDDED
expression
for
SINGLE
POLES
the in-place
soil
165
spring
stiffness
(Khs):
(3) where a and 8 are constants. surface to the point of interest. A revised expression for direct embedment foundations mathematically combining the series with the in-place soil the foundation
and D is the depth below
the ground
the lateral subgrade modulus (kh) for for use in Equation 1 was obtained by expression for the annulus spring in spring and dividing by the diameter of
(Bo)'
(4)
where
a = 5.7
and
8 = 0.40.
A similar analytical procedure was conducted to produce a revised subgrade modulus value for the vertical side shear moment spring. The vertical side shear moment spring was considered to consist of two vertical force springs connected in series by a rigid link; one spring represented the annulus stiffness and the second spring represented the in-place soil, with both springs considered to be elasticperfectly plastic. Again considering the two limiting conditions such that E =E and E »E and combining the two springs in series resulted a ~ a s in the following expressions for the annulus stiffness (Ke) and the in-place soil stiffness (Ke) : a -s 0.55
E
a
B2
(5)
Kea and
(B/B o )2 -1
(6)
Mathematically
combining
these
expressions
in series
and
rearranging
to obtain a subgrade modulus (ke) for the combined vertical shear moment spring resulted in the following expression: Bo (B/B
0/
side
(7)
+ (E a /E s ) - 1 where Ea' Es' Band Bo' are as defined previously. For the condition where E s is greater than E a , the expressions kh and ke reduce to corresponding subgrade modulus values for annulus backfill as the Es to Ea ratio approaches infinity.
for the
TRANSMISSION LINE TOWERS FOUNDATIONS
166
Ultimate
Capacity.--For
direct
embedment
foundations,
the
computation of the ultimate capacity (lateral pressure), Pult' of the lateral spring must consider several potential conditions; 1) the failure mechanism may be contained within the interior of the annulus (e.g., when the annulus material is much weaker than the in-situ soil), 2) the annulus material may act as part of the foundation and the failure mechanism will be located exclusively in the in-situ soil (e.g. when an annulus backfill such as concrete is much stronger than the in-situ soil), and 3) the failure mechanism involves both the annulus backfill and the in-situ soil. For the second condition, the foundation may effectively be designed as a drilled shaft foundation and Hansen's (5) solution used to determine Pult' In the case of the third condition, it is assumed that the percentage of the foundation failure mechanism (failure surface) contained within the annulus will be very small since the annulus thickness is generally on the order of less than 1 foot. Therefore, Hansen's equation may also be used to determine the ultimate lateral pressure using the strength properties of the in-situ soil and assuming the effective diameter of the foundation to be equal to the diameter of the embedded structure. An approximate solution for the ultimate lateral pressure based upon a failure mechanism contained wi thin the annulus (Condition 1) was developed based upon the simplified geometry shown in Figure 5. The circular cross-section of a direct embedment foundation and annulus were represented by concentric squares and a failure surface consisting of a series of rigid wedges was assumed. The expression obtained for the ultimate pressure was arranged in the form of:
P
(8)
ult
where q
is the effective
overburden
pressure
at a given depth in the
annulusmbackfill, ca is the cohesion of the annulus backfill, and KQm and Kcm are bearing capacity factors presented in Appendix A. THe bearing capacity factors were adjusted to provide the same numerical values for Pult as the Hansen was large.
solution
(5) when
the ratio of B to Bo
In the case of the vertical side shear moment spring, it was assumed that two potential failure surfaces must be considered due to the manner of construction of direct embedment foundations; 1) the interface between the foundation and the annulus material and 2) the interface between the annulus backfill and the in-situ soil. The development of expressions to determine the ultimate vertical side shear moment followed explicitly the formulation developed for the drilled shaft four-spring model (1). Appendix B summarizes the relationships for ultimate vertical side shear force (V z ) and ultimate side shear moment (Mzult)' The influence of construction method on the available shear strength at the two interface locations is accommodated by the inclusion of strength reduction factors a and a rs shown in Appendix B. ra
167
DIRECTLY EMBEDDED SINGLE POLES AT-REST
EARTH PRESSURE
ANNULUS m
BACKFILL
ASSUMED
RIGID BOUNDARY
IN-PLACE
NATURAL
SOil (A) CROSS-SECTION
OF FOUNDATION AT-REST
(B)
ASSUMED
FAILURE'
- ANNULUS PRESSURE
WEDGES
SYSTEM
RESULT ANT
AND FORCES
FIGURE 5.--Simplified Model for Failure Totally Within the Annulus
Surface Contained
The direct embedment foundation model and the original PADLL drilled shaft model are contained in a new EPRI computer program MFAD (Moment Foundation Analysis and Design). Thus, MFAD has d~sign/analysis capabilities for both drilled shaft and direct embedment foundations (6). Field Testing Program In order to obtain comprehensive data on the performance of direct embedment foundations subjected to high overturning moments, a series of 10 full-scale direct embedment foundation load tests were conducted at various test sites. Subsurface Investigation.--In order to characterize subsurface conditions and select stiffness and strength parameters for design and analysis of the test foundations, two borings were typically drilled at each test site. The initial boring at a each site included standard penetration testing, pocket penetrometer testing, and visual classification. Following the determination of the stratigraphy at each test site, a second boring was drilled in close proximity to the first. Pressuremeter tests were conducted at selected intervals and undisturbed
soil samples were extracted
for laboratory
testing.
In addition, samples of backfill materials were obtained prior to design of the test foundations for laboratory testing to obtain strength and deformation parameters which could be used in conjunction with the model to design the test foundations. The backfill material
J68
TRANSMISSION LINE TOWERS FOUNDATIONS
consisted of either compacted native augering of the foundation hole or select
soil excavated material (crushed
during stone).
the
Instrumentation.--Surface instrumentation, consisting of 6 dial gages, for the field load tests was installed at the ground-line to measure displacement and rotation of the foundation in the plane of and perpendicular to the direction of the applied loads. Survey measurements were made with a transit to determine the deflection of the top of the pole and to measure large ground-line movements. The below-ground ins trumentation consis ted of s train gages bonded to the steel or concrete poles at various intervals below the ground surface. The strain gages were used to determine the below ground bending moment distribution in the foundation. No strain gages were mounted on the one wood pole tested. Loading of all of cable at a convenient were applied to the truck. The applied mounted in series electronic load cell back-calculating the top of the pole).
the tests poles was accomplished by attaching a location near the top of the pole. Test loads cable by means of a winch mounted on a dozer or load was measured using either two dynamometers or a dynamometer mounted in series with an connected to the loading cable (as well as by applied load from the measured deflection of the
Foundation Test Design.--The full-scale test foundations were selected from available transmission poles owned by the utilities sponsoring the load tests. The embedment depths for the test foundations were computed using the design capabilities of the computer program MFAD for an applied ground-line moment equal the ultimate ground-line capacity of the transmission pole divided by a factor of safety of 1.5; so that geotechnical failure of the foundation would occur well before structural failure. The load testing program included 7 tubular steel poles, 2 prestressed coricrete poles, and one timber pole. The two concrete poles were embedded using native soil (silty clay) as backfill material and the remaining 8 load tests utilized various crushed stone backfills. The test poles varied from 65 to 115 feet in length, 27 to 38 inches in diameter, and the embedded lengths varied from 7.7 to 11.5 feet. In general, the backfill was well-compacted, with the exception of one test using native soil backfill and one test using select backfill in which the backfill was not compacted or only lightly tamped, respectively. The test loads were applied to the pole in increments keyed to percentages of the ultimate moment capacity of the foundation predicted by the model. Each load increment was maintained on the test foundation until the rate of ground-line deflection decreased to 0.01 inches/hour. Typically, three load-unload cycles were applied prior to reaching the predicted ultimate foundation capacity. Figure 6 shows a typical load test curve (applied ground-line moment vs. ground-line deflection) obtained from the testing program. The load tests were concluded when an applied load increment could not be sustainfd and large ground-line deflections occurred (the exception is Test No. 10 in which the applied moment was increased until the factor
DIRECTLY
EMBEDDED
169
SINGLE POLES
of safety on the structural capacity of the pole was reduced to approximately 1.1 without reaching a limiting geotechnical load). Consequently, the maximum applied moment was adopted as the ultimate capacity of the foundation. In the case of Test No. 10, the ultimate capacity was estimated by extrapolating the load-deflection curve toward a limiting value. Model Predictions Versus Field Load Test Data.--The primary purpose of the field testing program was to provide a data base for evaluation of the predictive capabilities of the direct embedment foundation model with respect to ultimate foundation capacity and the loaddeflection and load-rotation behavior at loads less than the ultimate capacity. Consequently, the computer program MFAD the foundations and, thereby, also provided a foundation's performance prior to the load tests. made to the predictions subsequent to the tests, account for the as-constructed augered hole sizes place density of the compacted annulus backfill. Figure 7 provides
a graphical
comparison
was used to design prediction of the Adjustments were as appropriate, to and the actual in-
of the predicted
ultimate
moment (Mult) capacity versus the maximum applied ground-line moment (Mmax) for the 10 test foundations. The ratio of M lt to Mmax ranged from 1.04 to 0.64 with an average value equal to O.~l. Therefore, in general, the model tended to underpredict the ultimate geotechnical capacity of the foundations by approximately 20 percent on the average. A comparison was also made of the applied (Ma) versus predicted ground-line moment (M) values obtained from moment-deflection and moment-rotation curvef developed from the load test results and computer predictions, respectively. Figure 8 presents a graphical comparison of M and M for data points taken at 0.5, 1.0, and 2.0 inches of defle~tion aKd Figure 9 presents a similar plot for data points taken at a .5, 1.0~ and 2.0 degrees of rotation for all of the load tests (except Tests 1 and 4 which had very loose backfill and, thus, were not considered in the deflection/rotation data base). In the case of defl~ction, the mean value of Mp/Ma' the standard deviation and coefficient of variation of M /Ma equal 1.16, 0.16 and 18.6 percent, respectively. The correspondi~g values for the mean of Mp/Ma' the standard deviation and coefficient of variation rotation data are 1.08, 0.15, and 15.6 percent, respectively. Summary
for
the
and Conclusions
A semi-empirical model for computing the ultimate lateral load capacity and load-displacement response of direct embedment foundations was presented. Comparisons of load test results with model predictions indicate that the model conservatively underpredicts ultimate moment capacity by approximately 20 percent. Comparisons of the ratio of predicted moment to applied moment for deflection/rotation at 0.5, 1.0 and 2.0 inches/degrees indicate good correlation. For deflection, the mean value of M /Ma, the standard deviation, and coefficient of variation are ~.16, 0.16 and 18.6 percent,
0
170
TRANSMISSION LINE TOWERS FOUNDATIONS
,... 1600
Iu. I
2S
---------------
I-
(1189) MAXIMUM APPLIED MOMENT
r5 1200
~ o ~
(1060) MULT (MFAD PREDICTION)
w
z
:iI
800
o
Z
::>
o ct CJ
400
o W ..J
Cl. Cl.
«
4
2
6
DEFLECTION
FIGURE 6.--Typical
-'
zI ~
n
"-,...
'" -;:
~
<:
•• Y
(5 f::. () 2000 BOO C --' 1600 I£l. 1200 I 9534ID0, It) W ::; ~ 0:;)w:lI-I::; LEGEND: 01 est I1 LINE 62 B
::E
8 AT GROUND-LINE
10
Applied Moment/Deflection
7
/
12
I
Curve
(Test No.3)
'Jc
OF ~QUALITY
o
o
400
BOO
1200
14
(IN)
1600
2000
MAXIMUM APPLIED MOMENT, MMAx, (K-FT)
FIGURE 7.--Predicted Ultimate Capacity vs. Maximum Applied Moment
16
DIRECTLY EMBEDDED SINGLE POLES
171
"00
'••0
••.• 1
u.
DEFLECTION
0.S:1.0:AND
2.0'
too
I
LINE OF
>::
L
~ ~
tao.
Z I.U
::<
o ao. ~ a ~ () II' a I.U
LEGEND •
I.U
a:
o
0..
<)
TEST
•
'00
()
<)
~ ~ •
200
200
.00
100
100
2
• • •
.. T
10
1000
1200
1 COO
1100
APPLIED MOMENT,M.(K-FT)
FIGURE 8.--Applied vs. Predicted Moment at Ground-Line Deflections of 0.5, 1.0 and 2.0 Inches "'0 I ROTATlON'0.S:1.0:AND
2.0'
140'
t.I,
LINE OF EQUALITY
~ U. I
:: "0' L
~ ~ z
W
10'
::<
o
::<
~ 0..
a
co. '00
a: () a I.U
LEGEND • TEar
200
o <)
•
<00
I ••
100
•
o
I 1
&
•
•
10
1000
1200
~
100
2
•
.
1COO
,.00
APPLIED MOMENT.M.(K-FT)
FIGURE 9.--Applied vs. Predicted Moment at Ground-Line Rotations of 0.5, 1.0 and 2.0 Degrees
TRANSMISSION LINE TOWERS FOUNDATIONS
172
respectively. For rotation the corresponding and 15.6 percent, respectively.
values
are
1.08,
0.15
Acknowledgments The research described herein was cosponsored by the Electric Power Research Institute, Palo Alto, California (Project RP 1280-3), Empire State Electric Energy Research Corporation (Project 85-33), Delmarva Power Company, Jersey Central Power & Light Company, New York State Electric & Gas, Pennsylvania Power & Light Company, Potomac Electric Company, Virginia Electric Power Company, Kansas Gas and Electric Company, and Public Service Electric & Gas Company. References GAl Consultants, Inc., "Laterally Loaded Drilled Pier Research," Volumes I and II, Electric Power Research Institute Report EL-2197, Project 1280-1, Palo Alto, California, January 1982. 2.
DiGioia, A.M., Davidson, H.L., and Donovan, T.D., "Laterally Loaded Drilled Piers, A Design Model," Proceedings of Drilled Piers and Caissons Session, ASCE National Convention, St. Louis, Missouri, October 28, 1981, pp. 132-149.
3.
Reese, L.C., and Welch, R., "Lateral Loading of Deep Foundations in Stiff Clay," Journal of Geotechnical Engineering Division, ASCE, Vol. 101, No. GT7, July 1975, pp. 633-649.
4.
Ivey, D.1., "Theory, Resistance of a Drilled Shaft Footing to Overturning Loads," Texas Transportation Institute, Research Report No. 105-1, February 1968.
5.
Hansen, J. Brinch, "The Ultimate Resistance of Rigid Piles Against Transversal Forces," The Danish Geotechnical Institute Bulletin, No. 12, 1961, pp. 5-9.
6.
Final Report, EPRI Research Project 1280-3, Volume II, "Research Documentation," Electric Power Research Institute, Palo Alto, CA.
DIRECiL Y EMBEDDED SINGLE POLES
173
Appendix A - Bearing Capacity Factors for Approximate Ultimate Lateral Annulus Bearing Pressure The earth pressure coefficients for overburden cohesion (Kcm ) are determined as follows: K
Solution
pressure
to
(K ) qm
and
+ 2 Tan
qm
where: = Angle of internal
¢a
K om = coefficient 1m F
of at-rest earth pressure = 1-sin
= Bearing (Sin S (Cos S + Tan
=
friction for the annulus backfill
- Tan
-0.652 o when Aq
(Tan a - Tan
Correction Factor for Overburden
Capacity
Pressure
Term
o
a
=
0 -1
2
+ 0.230 0 x 10-2
=
F 2m
__ a
Tan a+ _a_n_a + a. Tan a+1 Tan S (Cas T (Sin S S + Tan Tan
(l+Tana-Tan
x
(l + Tan
Bearing Capacity
Correction
Factor for Cohesion Term
-2 1.233 + 0.103
-4 - 0.655 x 10
0.196 x 10
3
4
+ 0.801
a Tan -1(B~
-
7~
< a45 0 +\2) (Oal
o
45 when (~o)
45
o
o
(45
Tan
~ \
Tan (45
+
0
Bo)<[Tan
B
(45
2)
+ 2)/Tan
(45
2)]
+ 1
o
1 + Tan (45 1[Tan ~(45 When(Bo» + ITan
or~
(B/Bo) (2 when 0B B \
2)] +
+ 1
2) and
when
-.) .J::>.
Appendix
B - Expressions
to Determine
the Ultimate
Definition F =
n
v =
t
Foundation-Annulus
Resultant Normal Force on Foundatoin Perimeter Resultant Horizontal Shear Stresses on Foundation Perimeter
B
2!.
4
a
rmax
B
cr
0
---
~a
4
Side Shear
Force (per unit length of foundation) Mzult'"
(11
+ -
where:
1
ra
a ra B 0(ca +
ca Cs ¢a
~s a ra a
rs
Tan 'a)
rmax
(C1l 1 B--s_+_a rs \ 4 3
P ult
-(~
4 rs
....., ;:::0
>-
rmax Tan ")
B0
Z C/)
~
C/) C/)
c)s
(5
(~B)+(tars
l
B Tan ~s)
Z
rmax Tan ~ a)
a rs B(C s +~
rmax Tan
~s)
tTJ
o-3 :<
o
11
Eccentricity o
a
c 0~ a
ia
---
B
B
rmax
B or Tan;\ a )
ara B 2 ( 4 ca
B
cr
3
Ultimate Side Shear Moment (per unit length of foundation) of Force Vz
Soil Interface
Z
-2-
M
x
Foundations
Annulus-Natural
-a 4
11
--;-a ult 4
Maximum Normal Stress Acting on Foundation Perimeter Ultimate Vertical
Interface
rmax
(~B o) +(ta ra v z=
for Direct Embedment
1IB
0
ra P
cr
Side Shear Moment
+
tTJ ;:::0
2 -a 3 rmax Tan ~a)
zult
V
Diameter of Foundation Outside Diameter of Annulus Cohesion of Annulus Material Cohesion of Natural Soil Angle of Internal Friction of Angle of Internal Friction of Strength Reduction Factor for Strength Reduction Factor for
z
ar s
s Z B 2 \4 c
rmax
s Tan ~)
C/)
zult
oc::
V
t:i
M
z
f-1I
+ %a
"I1
Z >-
.....,
(5
Z C/) Annulus Material Natural Soil Annulus-foundation Interface Annulus-Natural Soil Interface
Horizontally Jean-Louis
Loaded Piles Next to a Trench
Briaud*, M.ASCE, Larry M. Tucker*, A.M.ASCE
Abstract The problem of a single pile subjected to a monotonic horizontal load next to a trench is addressed. In a first part a total of 12 pressuremeter tests are performed increasingly closer to a deep trench in clay and then in sand. The results show the influence of the trench at small strains and at large strains. In a second part a FEM analysis is performed in order to extend the PMT results to the case of variable trench depth. In a third part a method is proposed to modify the P-y curve to include the presence of a trench. Background There are a number of solutions to the problem of horizontally loaded piles (Baguelin et al., 1978; Briaud and Tucker, 1985; Broms, 1965; GAl Consultants, 1982; Menard et al., 1969; O'Neill and Gazioglu, 1984; O'Neill and Murchison, 1983; Poulos, 1971; Reese and Desai, 1977). These solutions do not address the case where a trench has to be opened near the pile (Figure 1). This article considers this case and gives recommendations for predicting the response of the pile. Previous studies on this particular problem include the work of Poulos (1978), Kratena et al. (1976) and Karcher (1980). Only Poulos' work is published in-English. Poulos considers the pile as a long vertical strip footing loaded horizontally in an elastic soil. The influence of the trench is taken into account by considering that the response of an element of this strip footing located at a horizontal distance x from the trench acts as a plate buried x deep into an elastic half space and pulled towards the surface. The limiting pressure for this elastic response is taken from the work of Meyerhof and Adams (1968) on uplift capacity of foundations. A series of model pile tests are also conducted. In this study, a series of pressuremeter tests were performed next to a trench in clay and next to a trench in sand; also a finite element simulation was used. The results of the field work and of the computer work are integrated to propose a method of prediction. The Sites and the Soils The (Briaud
two sites are located near the Te~as A&M University campus and Terry, 1985). The clay site consists of a very stiff
*Professor, Research Texas A&M University,
Associate, Department of Civil College Station, TX 77843, USA.
175
Engineering,
TRANSMISSION LINE TOWERS FOUNDATIONS
176
.-
QAY
.~
•
Q
4.5 8
I
,
.--
1.58-.13.5 B~ --./
D
I I
L
Tlf
B
.-
-.
V8
_p/
.48 8
cross ~CiI Oi
FIG.2.- Location of Tests.
J"~:,',
I' , , " ' '~'.~:,' ,
sl
H
~
r-:x:
7-~~
j ~
.~
~ , ~ e~
t: > <
11f
;.-
HIE = 5.5
Po.
,
"I'
'~.'~
H
'
HIE =9
11f
HIE = 3.5
[
HIB
= 2.5
M'15
2
O~""I""I""I""~ 10
o
RELATIVE
20
30
,(0
J
Fffi
-1
lj
~
~
H1B=4
t
_~ 1
jr
1:..
M HIE ,1,' - 2.:>
~~
0
INCREASE IN PROSE RAOIUS. dR/Ro (%) RELATIVE
FIG.3.- PMT Tests in Clay.
.J
~
~
3r
-i
I 1
•...
HIE = 4.5
~
'1
.,
~ .J
,
--.J oJ
8
~
~
~
'
't". ~
o .(~
[I
~r
ffif
I
~
I
_r
-
5
~
1''~'," ~
J)
TROO!
P.'If
PI..Ni VI Ell
FIG.l.- The Problem.
1 FT
nr
I
TRENOi
98
1.5B
SCALE
68
I
B
cross ~CiI Oi
.--
2.5 8 -.
TIDOi
ffif
PI..Ni VI Ell
L-1. ~
~
3.58
TIDQj
5.5 B
H
1 FT
IO-
2.5 8 -
M
~
SCAl.£
10
20
I 30
INCREASE IN PROBE RAOI~
FIG.4.- PMT Tests in Sand.
.(0 dR/Ro
~ ...J J 50 (:)
HORIZONTALLY LOADED PILES
I i
plastic clay with the following average properties over the first 6.1 m (20 ft); plastic limit 21%, liquid limit 54%, water content 24%, unit weight 19.8 kN/m3 (126 lb/ft3), undrained shear strength from unconfined compression tests 114.9 kPa (1.2 tsf), electric cone penetrometer point resistance 1916 kPa (20 tsf). The water table is 5.2 m (17 ft) deep. The sand site consists of a medium dense fine silty sand with the following average properties over the first 20 feet: dry unit weight 17 kN/m3 (108 pcf), water content 12.9%, 15% passing sieve no. 200, friction angle from direct shear tests 31°, SPT blow count 18.5 blows per 30 cm (18.5 bpf). The water table is 7.3 m (24 ft) deep. The Pressuremeter
Tests
A series of pressuremeter tests were performed. The pressuremeter used was the TEXAM (Roctest, 1983); the probe diameter is 74 rom (2.91 in.) and the inflatable length of the probe is 49 cm 09.3 in.). The boreholes were prepared by using a hand auger and the middle of the probe was placed at a depth of 60 cm (2 ft). A trench was opened which was 0.91 m deep, 0.45 m wide, 1.83 m long (3 x 1.5 x 6 ft). The pressuremeter boreholes were drilled at various distances from the trench as shown on Figure 2. The test results are shown on Figures 3 and 4. These pressuremeter curves show the decrease in soil resistance as the pressuremeter gets closer to the trench. The shape of the pressuremeter curve is normal for tests far away from the trench but for tests performed close to the trench some curves show a peak. This peak is especially noticeable for the tests in sand. Therefore it is logical to conclude that the failure of a pile horizontally loaded near a trench would be more sudden in sand than in clay. From each pressuremeter curve, a modulus Eo and a limit pressure PI.. were calculated. When the pressuremeter curve displayed a peak, the peak value was used as the limit pressure. Figures 5 and 6 show the variation of Eo and PI.. as a function of the distance between the pressuremeter and the trench. Figure 7 is similar to Figures 5 and 6 except that the vertical scale is normalized. The results show similar trends in sand and in clay. However the influence of the trench is felt more severely and further away at the limit pressure than for the modulus. The crack pattern developing on the vertical face of the excavation as a result of the pressuremeter expansion was monitored. In the clay, cracks appeared only for the test closest to the trench (1.5 PMT diameter from the trench); a vertical and a horizontal crack appeared in the shape of a plus sign. In the sand, cracks developed for the tests at 1.5, 2.5, 4, and 6 PMT diameters from the trench. At 6 diameters one single vertical crack was apparent while at 1.5 diameters a 60 cm by 60 cm (2 x 2 ft) block of sand fell into the excavation. Note that if the sand had been clean and dry, the trench would not have stood up.
177
178
TRANSMISSION LINE TOWERS FOUNDA TrONS
.
0/ :17, nr CLAY LIMIT4 PRESSURE ~ SAt'ID
I/
:t
.-
/ I,I l2l610/10l•20° 15 20 5I° 5• 0._------/ L/ / 1/ -" • --- .- l20l 0/ /8L
4/' MJIUlJS
FIG.S.-
o
B B
H
0H
.:------ / /
• .-0- -
HIE
op
OJ
P'M)!XJLUS
FIG.6.- Eo,P1 vs.R/B
Eo and PI vs. RIB (Sand)
(Clay)
1.0
_ -0 0.8
0.6
0.4 B
H
lfU
0.2
o o
8
6
2
l2
10
HIE
FIG.7.- Normalized
Parameters
as a Function
of RIB.
FIXED
x
""T1
PLN£
u...
I F'l..NE A'W..YSIS IN FREE n£CPtIN I 0W: ';fin! LI DISPLAC2'8H I~ ~ 110
B
'""
Cf UD 11-iE PILE a4H
I~
N\ALYSIS
H
0.25_
ij --- ~-I\.
16 B 8B
Cf TI£ PILE DISPlA..-m.£NT
PrE
~ FIXED
(TREK.,)
FIG.8.- Mesh in Horizontal
Plane
'FIG.9.- Mesh in Vertical
Plane
179
HORIZONTALLY LOADED prLES FEM Simulation
in the Horizontal
Plane
The program CRACKTIP (1986) was used to simulate the problem in the horizontal plane. The soil was considered to be linear elastic; a typical mesh with boundary conditions is shown on Figure 8. This plane strain problem simulates the case of an infinitely long pile next to an infinitely deep excavation. The distance from the pile to the trench was varied from 1 to 10 pile diameters. In all cases, the pi Ie was pushed horizontally 6.35 rom (0.25 in.) towards the excavation and the stress Gi in the first element against the pile was recorded. In order to simulate the case where no trench exists a rigid boundary was placed at 10 pile diameters from the pile. The stress Gi in the case of no trench is called GNT' Figure 7 is a graph of Gi/GNT versus H/B, H being the distance from the pile to the trench and B the pile diameter. FEM Simulation
in the Vertical Plane
The same program was used to simulate the problem in the vertical plane. The soil was considered to be linear elastic; a typical mesh with boundary conditions is shown in Figure 9. This plane strain representation of the problem simulates the case of an infinitely long wall next to an infinitely long excavation. The distance from the pile to the trench as well as the depth of the trench were varied. In all cases, all the nodes of the pile were pushed horizontally 6.25 mm (0.25 in.) towards the trench and the stress Gi in the first element against the pile was recorded for each of the 8 layers of soil (Figure 9) including the case where no trench existed. The stress oi in the case of no trench is called GNT' Figures 10 to 17 summarize the results. In order to compare the results of the FEM analysis in the vertical plane and in the horizontal plane, the average of Gi/GNT for the vertical plane cases where the trench is as deep as the pile are plotted on Figure 7. The comparison shows that the wall loading (vertical plane) is generally more severe than the infinitely long pile (horizontal plane). Note however that the FEM wall loading case falls between the results for the modulus and for the limit pressure of the pressuremeter. Proposed
P-y Curve Approach
One of the most common ways of predicting the response of horizontally loaded piles is to use the concept of P-y curves (Reese and Desai, 1977). These P-y curves have been recommended first by Matlock and Reese and later by other authors (Briaud and Tucker, 1985; GAl Consultants, 1982; Menard et a1., 1969; O'Neill and Gazioglu, 1984; O'Neill and Murchison, 1983) these recommendations pertain to the case of a horizontal ground surface. When a trench exists at some distance from the pile the P-y curves are affected.
curve given
In order to include needs to be reduced y the soil reaction
the effect of a trench, the by a trench influence factor P will be less. The factor
P of the P-y since for a A will always A
180
A
i'i L :~
I4
/
OIL= 0.75 LI
I
D/l= O.SO D/l= I O.SO O/L=O~75 OIL= O.SO /
,and z
O,SO A =:'
0.75 0 O.SO
O~~
/
LINE TOWERS
¥P llT ¥
. /' 1
I6 6S 10 lro 124100 1~.J I0) 842III621.00 2,2
=.£1
ction OIL= 1.00 of
TRANSMISSION
>--
I
OIL= 0:75 /
j~ [
W,-l00
Dft_"
I
II
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FOUNDATIONS
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HORIZONT ALLY LOADED PILES be less than one and will represent the ratio of the soil reaction with trench over the soil reaction without trench. This factor A corresponds to the parameter plotted on the vertical axis of Figures 10 to 17. Therefore it is recommended that, in order to correct P-y curves for trench effect, P be multiplied by A obtained from Figures 10 to 17. The A values vary along the pile length and depend on the distance to the trench as well as the depth of the trench. I
I
Note that since these A values come from the elastic analysis of a horizontally loaded wall instead of a pile, they are conservative values. However they are conservative at small strains (elastic analysis) but not at large strains since, as shown on Figure 7, the wall analysis is between· the pressuremeter modulus curve and the limit pressure curve. It is also necessary to make a distinction between the case where 8. pile element is moving towards the trench and the case where it moves away from the trench. in other words the P-y curves need to be nonsymmetrical (Figure 18) with a reduced P-y curve towards the trench and an unreduced P-y curve away from the trench. This can be easily handled by a Beam Column program (Bogard and Matlock, 1977). In the case of a pile in sand which is within 6 diameters from a trench, Figure 6 shows that there is a need to use P-y curves which exhibit a peak. This peak occurs at a relative increase in cavity radius ilRc/Rc (Figure 6) of approximately 10%. It has been shown (Baguelin et a1., 1978; Briaud and Tucker, 1985) that this corresponds on the P-y curve to a y value equal to 0.10 Rpile. Therefore beyond 0.1 Rpile the P-y curves, in this special case, should be softened according to the shape of the pressuremeter curves on Figure 6. If this provision is not included in the P-y curves the pile response prediction will only be valid up to a displacement equal to 0.1 Rile. Alternatively the P-y curve can be obtained directly ~y performing pressuremeter tests at the site near the trench and using the method de~cribed by Briaud and Tucker (1985). Coaclasions A method is proposed to predict the response of piles loaded horizontally near a trench. In order to propose this method a series of pressuremeter tests were performed near a trench in sand and in c lay and a series of FEM simulations were conducted. The pressuremeter tests showed that: 1. When a deep trench is at5 pressuremeter diameters from the test the modulus is reduced to 80% of the modulus without trench and the limit pressure is reduced to 50%. A curve is presented to quantify the reduction as a function of the distance to the trench (Figure 7). 2. In sand pressuremeter tests within 6 pressuremeter diameters of the trench show a peak in the expansion curve. 3. In sand and in clay the modulus is less sensitive to the trench than the limit pressure. The FEM analysis shows the influence of the depth of the trench, a factor which was not investigated with the pressuremeter. The results allow to obtain the ·trench influence factor A . for various
181
182
TRANSMISSION
LINE TOWERS FOUNDATIONS
1.())
D/l = 1.()) 0.75
APPlIES
A =
TO
0.75 L
£.1 0.:0
TO
IPTH:
0.375 L
A"PliES
TO EE?TH:
0.875 L
TO
1.0 L
D/l = 1.())
U,...
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o
5
o
FIG.16
and
8
17.- Parameter
6
100
A as a Function
M
---.. Q~
~
p~:
p-y
FIG.18.-
P-y and
AP-y
Curves.
or ~
'
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z.
8
10
HORIZONTALLY LOADED PILES
183
depths of trench, distances between the pile and the trench, and depth along the pile (Figures 10 to 17). It is proposed to use this ;\ factor to reduce the P-y curves to;\ P-y curves in order to predict the response of piles loaded horizontally next to a trench. Full scale load tests need to be performed to evaluate the reliability of the proposed method. It must also be kept in mind that if the sand does not have a sufficient amount of fines and is either dry or submerged the trench alone is not stable. In all cases the stability of the trench alone must be established before addressing the problem of the horizontally loaded pile. Acknowledgments
This project was sponsored in part by Briaud Engineers. The following individuals participated in the project and are thanked for their contribution: Lopez, X., Gan, K.C., Chandra, D., Kon, C.J., Leonard, J.N., Pittenger, H.A., Schuller, R.E., and Webb, R.E. References
1. Baguelin, F., Jezequel, J.F., Shields, D.H., The Pressuremeter and Foundation Engineering, Transtech Publications, Rockport, Mass., 1978. 2. Bogard, D., Mat lock, H., "A Computer Program for the Analysis of Beam-Column under Static Axial and Lateral Loads," Offshore Technology Conference, Paper OTC 2953, 1977. 3. Briaud, J. L., Terry, T., "Texas A&M Uni versi ty Geotechnical Research Sites," Research Report, Civil Engineering, Texas A&M University, 1985. 4. Briaud, J.L., Tucker, L.M., "A Pressuremeter ~ethod for Laterally and Foundation EngiLoaded piles," Int. Con£. on Soil Mechanics neering, Vol. 3, p 1353, 1985. Journal of the 5. Broms, B.B., "Design of Laterally Loaded Piles," Soil Mechanics and Foundations Division, ASCE, Vol. 9 1 , SM3, 1965. 6. "CRACKTIP User's Manual," Civil Engineering, Texas A&MUniversity, 1986. 7. GAl Consultants, Inc., "Laterally Loaded Drilled pier Research: Volumne 1 and 2," Reports to EPRI, 1982. 8. Karcher, K., "Model Tests of the Bearing Capaci.ty of Horizontally Loaded piles on Slopes," Bautechnik 57, No. 10, pp 328-330,1980. 9. Kratena, J., Kysela, Z., Bartos, F., "A Model Study of the Interaction between Horizontally Loaded piles at the Crest of a Slope," Stravebnicky cas. 24, No.1, pp 44-52, 1976. 10. Menard, L., Bourdon, G., Gambi.n, M., Methode Generale de Calcul d'un Rideau ou pieu Sollicite Horizontalement en Fonction des Resultats pressiometriques," Sols-Soils No. 20/23, 1969. 11. Meyerhof, of G.G., Adams, J.1., "The Ultimate Uplift Capacity Foundations," Canadian Geot.echnical Journal, Vol. 5, No.4, pp 225-244, 1968. 12. O'Neill, M.W., Gazioglu, S.M., "An Evaluation of P-y Relationships in Clays," Research Report UHCE-84-3 to API, Civil Engineering, University of Houston, 1984.
184
TRANSMISSION LINE TOWERS FOUNDA TrONS
l3.-0'Neill, M.W., Murchison, J.M., "An Evaluation of P-y Relationships in Sands," Research Report GT-DF02-83 to API, Civil Engineering, University of Houston, 1983. 14. Poulos, H.G., "Behavior of Laterally Loaded piles: 1 - Single Piles," Journal of Soil Mechanics and Foundation Engineering, ASCE, Vol. 98, SM4, 1971. 15. Poulos, H.G., "Behavior of Laterally Loaded piles Near a Cut or Slope," Australian Geomechanics Journal, Vol. G6, No.1, 1978. 16. Reese, L.C., Desai, C.S., "Laterally Loaded Piles," Chapter 9 in Numerical Methods in Geotechnical Engineering, McGraw-Hill, 1977. 17. Roctest, Inc., "TEXAM Pressuremeter Operation Manual," Plattsburg, New York, 1983.
SUBJECT INDEX· Page number refers to first page of paper. Anchors, 57, 72, 81
Lattices, 15,39 Load tens, 128, 160
Bell footings, 110 Boring, 1
Marshes, 72, 81
Clays, 128, 175 Cone penetration tests, 39 Construction methods, 72, 81
Networks, 25
Drilled piers, 128 Drilled shafts, 142 Driven piles, 142
Pile foundations, 39 Piles, 175 Poles, 39, 160 Probabilistic methods, 1
Overconsolidated
Foundation design, 15,25, 72, 160 Foundation performance, 15 Framed structures, 15 France, 25
clays, 110
Sand, 57, 96, 128, 175 Shafts, 15 Site evaluation, 1,81 Soil investigations, 25 Soil suction, 110 Spread foundations, 96 Steel piles, 39 Subsurface investigations,
Granular materials, 142 Guyed towers, 15 Helixes, 72, 81 Horizontal loads, 175
Transmission Laboratory tests, 57 Lateral loads, 160
1,25, 128
towers, 25
Uplift resistance, 57, 96, 128, 142
185
AUTHOR INDEX Page number refers to first page of paper. O'Neill, Michael W., 110
Bragg, Richard A., 160 Briaud, Jean-Louis, 175
Pacal, Albert J., 128 Clemence, Samuel P., 72 Rodgers, Thomas E., Jr., 81 Das, Braja M., 57 DiGioia, Anthony M., Jr., 160
Sheikh, Shamim A., 110 Shively, Arthur W., 128 Spry, Mary J., 1
Filippas, Olga B., 1
Tedesco, Paul A., 15 Thomas, Walter G., 15 Trautmann, Charles H., 96 Tucker, Keith D., 142 Tucker, Larry M., 175
Gagneux, M., 25 Grigori, Mircea D., 1 Jin-Jaun, Yang, 57 Konstantinidis, Byron, 128 Kulhawy, Fred H., 1, 96
Verstraeten,
L,
Alexander J., 39
Weikart, Albert M., 72
Lapeyre, J. 25 Longo, Vito J., 160
Yazdanbod, Nicolaides, Costakis N., 96
186
Azaroghly, 110
