Precast Concrete Structures - Hubert Bachmann

Hubert Bachmann, Alfred Steinle Precast Concrete Structures

Hubert Bachmann / Alfred Steinle

Precast Concrete Structures

Dr.-Ing. Hubert Bachmann, Department Manager Ed. Zu¨blin AG Structural Engineering (TBK-S) Albstadtweg 3 D–70 567 Stuttgart Dr.-Ing. Alfred Steinle Alte Weinsteige 92 D–70 597 Stuttgart

Translated by Philip Thrift, Hannover/Germany Cover photo: The “Dancing Towers”, Hamburg/Germany; two office towers of 80m and 70m height, under construction. (Photo: Ed. Zu¨blin AG)

Bibliographic information published by the Deutsche Nationalbibliothek The Deutsche Nationalbibliothek lists this publication in the Deutsche Nationalbibliografie; detailed bibliographic data are available on the Internet at http://dnb.d-nb.de. c 2011 Wilhelm Ernst & Sohn, Verlag fu¨r Architektur und technische Wissenschaften GmbH & Co. KG, Rotherstr. 21, 10245 Berlin, Germany

All rights reserved, particularly those of translation into other languages. No part of this book may be reproduced in any form – by photocopy, microfilm or any other means – nor transmitted or translated into a machine language without permission in writing from the publisher. The reproduction of product descriptions, trade names and other designations in this book does not imply that these may be freely used by any person. These may be registered trade names or other designations protected by law even when they have not been specifically identified as such. All books published by Ernst & Sohn are carefully produced. Nevertheless, authors, editors and publisher accept no liability whatsoever for the accuracy of information contained in this or any book or for printing errors. Coverdesign: Sophie Bleifuß, Berlin Production: HillerMedien, Berlin Typesetting: Hagedorn Kommunikation, Viernheim Printing and binding: Betz-Druck, Darmstadt Printed in the Federal Republic of Germany. Printed on acid-free paper. ISBN 978-3-433-02960-2 Electronic version available. O-Book ISBN 978-3-433-60096-2

in memoriam Volker Hahn 10 April 1923 – 1 May 2009

The authors would like to dedicate this English edition to Prof. Dr.-Ing. Volker Hahn, their mentor and for many years their boss at Zu¨blin, and also co-author of the first German edition. Volker Hahn began his career at Ed. Zu¨blin AG in 1949 and in his role as development engineer established the main engineering office, which is still the technological heart of the company. He was one of the pioneers as computers were introduced into construction and with great farsightedness initiated important developments in precast concrete construction, transportation, specialist civil engineering, turnkey projects and environmental protection technology. He was a member of the Board of Directors at Ed. Zu¨blin AG from 1971 to 1987. It was in his capacity as board member that he made major contributions to the dynamic growth of the company, its ongoing economic success and its position as a technology leader. Young engineers were able to benefit from Volker Hahn’s superlative specialist knowledge through his lectures at the University of Stuttgart, where he was honorary professor. Zu¨blin House was built under his leadership, a project that lent new momentum to precast concrete construction.

Precast Concrete Structures. First Edition. Hubert Bachmann, Alfred Steinle c 2011 Ernst & Sohn GmbH & Co. KG. Published by Ernst & Sohn GmbH & Co. KG.

Preface Building with precast concrete components is as old as building with concrete itself. It was only in the second half of the last century, however, that this form of construction took on its industrialised form. Factors that contributed to this were, in particular, the development of heavy lifting equipment, the use of mechanised steel moulds and, more recently, automated manufacturing systems, for suspended floor elements especially. This work on precast concrete construction was first published in 1988 as part of the Beton-Kalender. A second version by the authors Alfred Steinle and Volker Hahn appeared in the same publication in 1995. These essays were turned into a book which was published in 1998 as part of the Bauingenieur-Praxis series of the Ernst & Sohn publishing house. A further treatise appeared in the 2009 edition of the Beton-Kalender, this time with Hubert Bachmann joining the original authors, and it was that version that became the second edition of the book in German. Inevitably, there have been some changes to the standards over the past 10 years. For example, the publication, following a long period of preparation, of the new DIN 1045 “Concrete, reinforced and prestressed concrete structures”. This standard has been approved by the building authorities for use in the Federal Republic of Germany since September 2002 and since 1 January 2005 is the only standard that may be used for concrete works. It was drawn up on the basis of the Euronorm EN 1992-1 “Design of concrete structures” (previously known as Eurocode 2) and therefore represents the translation of this Euronorm into national German practice. Furthermore, we are witnessing a fundamental change in the design of precast concrete components. The creation of the European Single Market led to the publication of the Construction Products Directive, which has been in force in Germany in the form of the Construction Products Act (Bauproduktengesetz) since 1992 and in the meantime has become part of the building regulations of the federal states which were revised to take account of this legislation. The directive renders it necessary to establish harmonised product standards specifically for the various precast concrete products so that in the end it will be possible to use all such components – labelled with the CE marking – throughout the European Union. With modern methods of construction making use of industrial methods of manufacture, which includes construction with factory-precast concrete components, the design of the individual elements, and also the entire structure, is heavily influenced by the factory production. On the manufacturing side, the growing trend towards mechanisation and automation in production is evident. The development of high-performance concrete provides us with the chance of employing these for precast concrete construction in particular because factory production presents excellent conditions for their use. For example, the first precast concrete compoPrecast Concrete Structures. First Edition. Hubert Bachmann, Alfred Steinle c 2011 Ernst & Sohn GmbH & Co. KG. Published by Ernst & Sohn GmbH & Co. KG.

VIII

Preface

nents made from ultra-high-strength concrete for bridges and façades have already been produced. Besides the industrial production of batches and series of components, we are seeing more and more one-offs being produced, which take advantage of the excellent production options in order to achieve a high standard of quality. These tendencies will become even more obvious as more and more progress is made in the development of concrete as a building material. The authors’ aim in writing this book is to map out the boundary conditions of factory prefabrication for architects and structural engineers and also to demonstrate the opportunities presented by this method of construction – in the expectation that this will contribute to the ongoing development of precast concrete structures. Stuttgart, November 2010 Ed. Zu¨blin AG

A. Steinle

H. Bachmann

Authors Alfred Steinle (b. 1936) turned Hahn’s lecture notes into a manuscript in the early 1970s, which then became the starting point for this book. After a number of years in bridgebuilding, Alfred Steinle also became heavily involved in precast concrete construction at Zu¨blin. His theoretical work covered bridge-building with torsion and section deformations in box-girder bridges and in precast concrete structures within the scope of the 6M system with corbels, notched beam ends and pocket foundations. In addition, he was a key figure in many precast concrete projects such as the 6M schools, the University of Riyadh, schools with foamed concrete wall panels in Iraq, Zu¨blin House and the construction of a modern automated precasting plant. Alfred Steinle retired in 1999 and by that time he had risen to the post of authorised signatory in the engineering office at Zu¨blin headquarters. Hubert Bachmann (b. 1959) began his career in 1976 with a training course on concrete and precast concrete construction in a precasting plant. After studying construction engineering and completing his doctorate at the University of Karlsruhe, he accepted a post in the structural engineering office of Ed. Zu¨blin AG, where he has worked since 1993. His duties include the detailed design of structures of all kinds plus research and development in the civil and structural engineering sectors. He has been presenting the series of Hahn lectures at the University of Stuttgart on the subject of the prefabrication of concrete components since 2003.

The authors were or are also intensively involved in construction industry associations, numerous technical bodies plus national and international standards committees dealing with precast concrete construction.


Contents Preface . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . VII Authors . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . IX Preliminary remarks . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1 Standards, leaflets and directives . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2 1

General . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5

1.1 1.2 1.3

The advantages of factory production . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5 Historical development . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6 European standardisation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9

2

Design of precast concrete structures . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15

2.1 2.1.1 2.1.2 2.1.3 2.1.4 2.2 2.2.1 2.2.2 2.2.3 2.2.4 2.2.5 2.2.6 2.2.7 2.3 2.3.1 2.3.2 2.3.3 2.3.4 2.3.5 2.4 2.4.1 2.4.2 2.4.3 2.4.4 2.4.5 2.5 2.6 2.6.1

Boundary conditions for precast concrete design . . . . . . . . . . . . . . . . . . . . . . . . . . . . 16 Production process . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 16 Tolerances . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 16 Transport and erection . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 21 Fire protection . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 23 Stability of precast concrete structures . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 28 Arrangement of stability elements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 28 Loads on stability elements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 33 Distribution of horizontal loads . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 45 Verification of building stability . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 58 Structural design of floor diaphragms . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 62 Structural design of vertical stability elements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 67 Design of perimeter ties to DIN 1045-1 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 72 Loadbearing elements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 78 Suspended floor elements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 78 Floor and roof beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 85 Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 94 Walls . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 97 Foundations . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 99 Precast concrete façades . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 101 Environmental influences and the requirements of building physics . . . . . . . 102 Façade design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 104 Joint design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 112 Façade fixings . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 116 Architectural façades . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 125 Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 130 Current design issues . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 139 Additions to cross-sections, floors with concrete topping . . . . . . . . . . . . . . . . . . 139


XII

Contents

2.6.2 2.6.3 2.6.4 2.6.5

Corbels and notched beam ends . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Lateral buckling . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Pad foundations . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Design for fire . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .

143 156 163 167

3

Joints between precast concrete elements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 175

3.1 3.1.1 3.1.2 3.1.3 3.1.4 3.2 3.2.1 3.2.2 3.2.3 3.2.4 3.2.5 3.2.6 3.3 3.3.1 3.3.2 3.3.3

Compression joints . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Butt joints . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Zones of support to DIN 1045-1 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Elastomeric bearings to DIN 4141 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Elastomeric bearings to DIN EN 1337 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Tension joints . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Welded joints . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Anchoring steel plates, dowels, studs and cast-in channels . . . . . . . . . . . . . . . . Shear dowels . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Screw couplers . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Transport fixings . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Retrofitted corbels . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Shear joints . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . General . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Floor diaphragms and wall plates – in-plane shear forces . . . . . . . . . . . . . . . . . . Joints in suspended floor slabs – out-of-plane shear forces . . . . . . . . . . . . . . . .

4

Factory production . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 213

4.1 4.2 4.2.1 4.2.2 4.2.3 4.2.4 4.2.5 4.3 4.3.1 4.3.2 4.3.3 4.4 4.4.1 4.4.2 4.5

Production methods . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Types of concrete in precast concrete construction . . . . . . . . . . . . . . . . . . . . . . . . Processing properties . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Self-compacting concrete (SCC) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Fibre-reinforced concrete . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Coloured and structured concrete surfaces . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Producing the concrete in the factory . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Heat treatment and curing . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Working hardened concrete surfaces . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Coating and cladding . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Installing the reinforcement in the factory . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Round bars and meshes . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Prestressing beds . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Quality control . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .

175 175 180 181 188 190 190 193 195 198 199 201 203 203 204 209

213 219 220 220 223 224 225 227 227 229 231 233 233 237 242

References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 245 Index . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 259

Preliminary remarks Chapter 1 contains general information about precast concrete construction, its history and the status of the relevant Euronorms. Chapter 2 explains the design of structures based on precast concrete elements and the design of the precast concrete elements themselves. Chapter 3 deals with joints. And to conclude this book, chapter 4 takes a look at the actual manufacture of precast concrete elements so that the reader gains a full understanding of this form of construction and can take into account the needs and intricacies of production. All this is seen from the viewpoint of the German construction industry. But with a view to the European Single Market and the activities of German companies abroad, the status of precast concrete construction in other countries is also considered to a certain extent. The authors have confined themselves in the main to structures in general. However, the fact that precast concrete construction has been able to secure sizeable market shares in many other areas of construction through the development of economic bespoke solutions should not go unmentioned. The following are just some areas where precast concrete construction has had considerable impact: x x x x x x x x x x

Bridges Tunnelling (tunnel segments) Pipes, pipe bridges, towers, masts, piles Detached houses Prefabricated basements, retaining walls Room modules, prefabricated garages Noise barriers Railway sleepers, slab tracks, guided bus tracks Agricultural structures Cooling tower trickle fill structures

The reader is referred to the specialist literature dealing with these specialist areas. This book also only describes structural or architectural precast concrete elements for buildings and structures and not “concrete products”, i.e. small-format components manufactured and stocked in great numbers and available from trade outlets, e.g. sewage pipes, paving stones, etc. The list of references has been extended since the previous edition. The references in the text have been retained on the whole because they contain potential solutions to fundamental problems that are still relevant today. Earlier articles on the theme of precast concrete construction worth consulting are those in the Beton-Kalender [1–3]. By the same token, reference to the general literature on reinforced concrete construction has been omitted and the reader is referred to the corresponding articles in the Beton-Kalender, unless they concern areas that also touch on the specific problems of precast concrete construction. Readers who wish to obtain a compressive overview of this subject are recommended to consult Koncz’s three-volume work dating from the 1960s [4] and the brochures published by the Fachvereinigung Precast Concrete Structures. First Edition. Hubert Bachmann, Alfred Steinle c 2011 Ernst & Sohn GmbH & Co. KG. Published by Ernst & Sohn GmbH & Co. KG.

2

Preliminary remarks

Deutscher Betonfertigteilbau e.V. [5–8]. As well as covering small-format concrete products, the Beton- und Fertigteil-Jahrbuch [9], which is published annually, also contains articles on various structural and architectural aspects of precast concrete construction, different in each edition. Comprehensive information on mass-produced concrete products can be found in [12], and a number of fundamental and general thoughts on industrialised building methods using precast concrete elements are to be found in [10, 11]. The books [13–16] are based on the lecture notes of a number of university professors. The DIN standards most relevant to this subject, in the editions on which this publication is based, are listed below. Also listed here are the directives of the Deutscher Ausschuss fu¨r Stahlbeton relevant to precast concrete construction and the leaflets published by the Deutscher Beton- und Bautechnik-Verein e.V. and the Fachvereinigung Deutscher Betonfertigteilbau e.V. Section 1.3 deals in more detail with the status of the development of Euronorms. Directives or leaflets containing further information are referenced separately in the text. Standards, leaflets and directives Table 1 DIN standards of NA 005 (NABau, Building & Civil Engineering Standards Committee) relevant to precast concrete construction (many available in English) DIN

Edition

Parts/Title

488 1045 1048 1055 1164 EN ISO 17660 4102

2009 2008 1991 2002-2007 2003-2005 2006-2007 1977-2004

4108 4109 4141 EN 1337 4149

1981-2009 2003-2006 1984-2008 2005 2005

4212

1986

4213

2003

4223

2003

Parts 1–7 Reinforcing steels Parts 1–4 Concrete, reinforced and prestressed concrete structures Parts 1–5 Testing concrete Parts 1–10 & 100 Actions on structures Parts 10–12 Special cement Parts 1 & 2 Welding – Welding of reinforcing steel Parts 1–4 & 22 Fire behaviour of building materials and building components Parts 1–10 Thermal insulation in buildings Parts 1 & 11 Sound insulation in buildings Parts 1–3 & 13 Structural bearings Part 3 Structural bearings – Elastomeric bearings Buildings in German earthquake areas – Design loads, analysis and structural design of buildings Reinforced concrete and prestressed concrete craneways; design and construction Application in structures of prefabricated reinforced components of lightweight aggregate concrete with open structure Parts 1–5 Prefabricated reinforced compliments of autoclaved aerated concrete Parts 100–103 (draft) Application of prefabricated reinforced components of autoclaved aerated concrete Part 100 Aggregates for concrete and mortar – Recycled aggregates Parts 2–5 Approval testing of welders – Fusion welding Parts 1–5 Stainless steels Concrete windows – Dimensioning, requirements, tests Stairs in buildings – Terminology, measuring rules, main dimensions Lightweight concrete wallboards – unreinforced

2008 4226 EN ISO 9606 EN 10088 18 057 18 065 18 162

2002 1999-2005 2005-2009 2005 2000 2000

3

Preliminary remarks

DIN

Edition

Parts/Title

18 200

2000

18 202 18 203

2005 1997

18 230 18 500

1998-2002 2006

18 515 18 516 18 540 18 542

1993-1998 1990-2009 2006 2009

18 800 18 801

2008 1983

Assessment of conformity for construction products – Certification of construction products by certification body Tolerances in building construction – Buildings Part 1 Tolerances in building construction – Prefabricated components made of concrete, reinforced concrete and prestressed concrete Parts 1–3 Structural fire protection in industrial buildings (pre-standard) Cast stones – Terminology, requirements, testing, inspection Parts 1 & 2 Cladding for external walls Parts 1 & 3–5 Cladding for external walls, ventilated at rear Sealing of exterior wall joints in buildings using joint sealings Sealing of outside wall joints with impregnated sealing tapes made of cellular plastics – Impregnated sealing tapes – Requirements and testing Parts 1–4 Steel structures Structural steel in building; design and construction

Table 2 DBV leaflets and status reports (Deutscher Beton- und Bautechnik-Verein e.V., German Concrete & Building Technology Association) (some available in English in the DBV’s “Concrete Best Practice” publication) Edition

Title Building technology

2005 2006 2006 2002

Multi-storey and basement car parks Structural carcass/building services interfaces – 2 parts Limiting cracking in reinforced and prestressed concrete Concrete cover and reinforcement Concrete technology

2001 2002 2004 2004 1996 2007

Steel fibre-reinforced concrete High-strength concrete Self-compacting concrete Concrete surface – concrete boundary zone Unformed concrete surfaces Special methods for testing wet concrete Construction works

2004

Fair-face concrete

2004 2006

Avoiding problems in placing concrete Concrete formwork Construction products

2002 2008 1996 1997 1999

Spacers and chairs for reinforcement Bending back reinforcing bars and requirements for bar casings Sealing materials for joints in buildings Release agents for concrete – part A: advice for selection and use Release agents for concrete – part B: tests Construction works in existing buildings

2008 2008 2008

Guidelines Fire protection Concrete and reinforcing steel

4

Preliminary remarks

Table 3 FDB leaflets (Fachvereinigung Deutscher Betonfertigteilbau e.V., German Precast Concrete Construction Association) (No. 1 available in English, all others in German only) No.

Edition

Title

1

2005

2

2005

3 4 5 6 7

2007 2006 2005 2006 2008

Fair-face concrete surfaces (surface appearance) of precast elements made of concrete and reinforced concrete Corrosion protection for inaccessible steel connecting elements (cast-in parts) in precast concrete components Design of precast concrete façades Fixing methods for precast concrete façades Checklist for precast concrete component drawings Fit calculations and tolerances for cast-in parts and connecting elements Fire protection requirements for precast concrete components

Table 4 DAfStb directives (Deutscher Ausschuss fu¨r Stahlbeton, German Reinforced Concrete Committee) (available in German only) Edition

Title

1989 1995

Heat treatment of concrete Production of concrete using residual mixing water as well as concrete and mortar residues Loading tests on solid structures Protection of and repairs to concrete components (parts 1–4) Self-compacting concrete (SCC directive) Concrete construction in connection with substances hazardous to water Concrete to EN 206-1 and DIN 1045-2 with recycled aggregates to DIN 4226-100 Concrete with prolonged working time (retarded concrete) Production and use of cement-bonded grouts Measures to prevent damaging alkaline reactions in concrete (alkali directive), parts 1–3

2000 2001 2003 2004 2004 2006 2006 2007

1

General

1.1

The advantages of factory production

The corporate goal behind the use of a production method that is to establish itself in the marketplace must be: to produce a product better or cheaper or faster than the competition. The optimum situation would be if each “or” could be replaced by “and”. So what is the situation with construction using precast concrete elements? a) Improved quality x

x

x

x

Production in an indoor environment results in better working conditions with correspondingly better productivity than would be the case on a building site, and that has an effect on quality, too. In the factory production situation, training makes it easier to compensate for the ongoing severe shortage of skilled workers in the construction industry. Steel moulds can be used for standard elements or large batches, which enables a high degree of dimensional accuracy to be attained. Factory production enables a specific concrete quality to be achieved. Only through factory production is it possible to produce concrete components with architectural textures and colours, especially for façade designs. As with other branches of industry outside the construction sector, factory production results in more efficient quality management.

b) Lower production costs x

x

x

x

x

The main purpose of precast concrete construction is to reduce the cost of the formwork. Several components can be produced in the same formwork, i.e. mould. And of course, large batches are advantageous. Although mould types suited to the method of production (e.g. rigid moulds with few fold-down parts) demand a design approach that suits the production, this does lead to high mould reuses. Another reason for precast concrete construction was undoubtedly the reduction or total elimination of scaffolding costs. Factory production enables the use of mechanisation and automation, which in turn can result in a substantial reduction in the number of working hours necessary. However, if a factory’s capacity is not fully exploited, this can be a disadvantage because of the ensuing high proportion of fixed costs. Material savings arise from the possibility of using thin component cross-sections corresponding to the structural requirements, i.e. double-T or T-sections instead of rectangular sections. The advantage of the (possibly) lower weight of the concrete is in many cases only made possible through the higher quality of the concrete due to the factory production methods. One typical example of saving material and weight is resolving a solid slab into a hollow-core slab. And this is only possible with precast concrete construction. Prestressing is easy to achieve in the form of pretensioning in the prestressing bed.


6

1 General

x

One considerable cost factor for a precasting factory is of course the cost of transport, which limits the radius of activities and consequently the potential market for a precasting plant and hence its size. This does not represent a hindrance for the precast concrete market as a whole because there are now proficient precasting works within economic reach of any location.

c) Faster construction x

x

x

x

One big advantage of precast concrete construction is the potential to shorten the construction time. For example, wall and suspended floor elements can be produced simultaneously, even while the foundations are still being built. Production, and to a large extent erection as well, can take place during the winter. No extensive, elaborate on-site facilities are required. The structural carcass is dry and ready for immediate loading immediately after erection. The financial savings associated with a shorter construction time and the chance of generating revenue at an earlier date are important, often underestimated, reasons for precast concrete construction, particularly for industrial buildings. However, it should not be forgotten that structures made from precast concrete components often require a higher planning and design input. On the other hand, this input can be substantially reduced by using a standardised precast component system. The first CAD applications in reinforced concrete construction originated in precast concrete.

1.2

Historical development

Prefabrication, i.e. the building of components remote from their intended location in the structure, followed by subsequent erection is a method of construction that is as old as building with reinforced concrete itself. However, the development of modern construction with precast reinforced concrete components from its origins to a form of industrialised building only took place over the past 60 years. Ref. [20] contains a detailed description of the development of prefabricated housing in Germany up to 1945. Even though we might not be able to designate the first reinforced concrete flower tubs or boats of Joseph Monier or Joseph Louis Lambot in the mid-19th century as prefabricated “components” (Fig. 1.1), the first serious trials with structural precast reinforced concrete components did take place around 1900 (e.g. Coignet’s casino building in Biarritz, France, in 1891, and the prefabricated railway signalman and gatekeeper lodges of Hennebique and Zu¨blin in 1896, Fig. 1.2) [17]. This development continued in the first half of the 20th century throughout Europe and the USA, albeit only tentatively. The main reason for this was the lack of larger and flexible lifting equipment during this period. The real breakthrough did not come until after the Second World War [18]. In a first phase from 1945 to 1960 it was the extraordinary demand for housing that presented the building industry with a huge challenge. During this period it was the French (e.g. Camus, Estiot) and Scandinavian (e.g. Larsson, Nielsen) systems that provided decisive momentum

7

1.2 Historical development

Fig. 1.2 Fig. 1.1

Prefabricated signalman’s lodge (c. 1900) [17]

Joseph Monier (c. 1850) [17]

for construction with large-format panels. Their patents – through licensees – also dominated the German market. In the second phase, about 1960 to 1973 (see also [18]), growing prosperity led to a rise in demand for owner-occupied housing with a higher standard of comfort. Inflationary tendencies resulted in a huge amount of investment in property, and the increasing shortage of skilled workers was another reason that forced production to be transferred to factories, which in turn helped precast concrete construction to achieve a breakthrough. Alongside housebuilding, the increased need for more schools, colleges and universities led to the establishment of fully developed loadbearing skeletal frame systems with columns, beams and long-span floors (7.20 m/8.40 m). Buildings for industry and sports centres resulted in standardised product ranges for single-storey sheds made from precast columns and prestressed rafters and purlins or sawtooth roofs. The third phase, about 1973 to 1985, was marked by a serious crisis for the German construction sector, first and foremost housebuilding. This was compensated for to a certain degree by increased demand in the oil-exporting countries. Housing, school, university and office building construction projects were carried out in those countries, which opened up completely new dimensions in the industrialisation of precast concrete structures. However, the fall in the price of oil led to this compensatory business almost drying up in the early 1980s. In the years after 1985 a general economic upswing resulted in colossal improvements for the construction sector as well. However, the high wage and social security costs forced precasting plants to switch to mechanised and automated methods of production. Since late 1989 Germany has seen renewed demand for housing to meet the needs of immigrants plus migrants from former East Germany. The opening of the border with the former German Democratic Republic in 1990 resulted in major challenges for the building industry in the ex-GDR.

8

1 General

Fig. 1.3 Concrete products and prefabricated elements in Germany: concrete products in total compared to large-format precast concrete elements (top); large-format prefabricated elements for structures (bottom)

New noise abatement legislation is one of the results of the growing awareness of environmental aspects, which has led to an increased demand for products such as noise barriers.

1.3 European standardisation

9

But the increased demand for building work after German unification was short lived. In the period from about 1994 to 2004, the construction sector experienced almost 10 years of decline, coupled with a drastic reduction in the number of employees and a rise in the number of insolvencies, even of large companies. This fact is also revealed by the production statistics for concrete goods and precast components, which are shown in Fig. 1.3. Fortunately, we have seen a change in fortunes since 2005. 1.3

European standardisation

The creation of the European Single Market has been accompanied by the vigorous development of European codes of practice. Most important here is the adoption of the “Construction Products Directive” (CPD) by the European Commission. This has been in force in Germany in the form of the Bauproduktengesetz (Construction Products Act) since 1992 and is crucial to the building industry. In the meantime, the building regulations of Germany’s federal states have been updated because the individual states will continue to be responsible for building regulations. The CPD defines “essential requirements” to be satisfied by “construction works” (and not just the construction products) in a general form. These are: 1. Mechanical resistance and stability 2. Safety in case of fire 3. Hygiene, health and the environment 4. Safety in use 5. Protection against noise 6. Energy saving and heat insulation. These requirements are concretised in six “base documents”, which are intended to form the foundation for “mandates” for preparing harmonised European standards (or directives for European approvals). These mandates must also include requirements for categories and performance classes for individual products (e.g. for static loads only, fire safety rating, etc.). The Euronorms (EN) then have to be drafted by the European Committee for Standardisation (CEN, based in Brussels). Products for which “conformity” with these harmonised Euronorms can be verified will in future be labelled with the CE marking (see also section 4.5). To date, the European Commission has issued the CEN with mandates for the standardisation of 30 product families. The standardisation work is carried out in so-called Technical Committees (TC) or Subcommittees (SC) and their associated Working Groups (WG) or Task Groups (TG). Once a CEN standard has been adopted by the “qualified” majority of the EEC and EFTA member states, all member states are obliged to adopt this standard, even if it was not “mandated” by the European Commission. In the case of “mandated” standardised Euronorms, no changes or additions are then possible when these are incorporated into the building legislation of Germany’s federal states (a situation that is different from the DIN standards in the past) because that would lead to new “trade barriers”.

10

1 General

Fig. 1.4 Systems for attestation of conformity procedures according to the Construction Products Directive (CPD) [29]

One key new development is that the supreme building authorities of the federal states have now published the “Construction Products Lists A, B and C” – standardised documents compiled by the Deutsches Institut fu¨r Bautechnik (DIBt, German Institute of Building Technology) [28]. Construction Products List A Part 1 contains construction products that have to comply with building authority requirements (e.g. suspended floor slabs, reinforcing steel, etc.). This corresponds to the building authority approval of the past.

11


Construction Products List A Part 2 contains construction products that require only a National Test Certificate (e.g. non-loadbearing lightweight partitions). Construction Products List B contains all those construction products that may be placed on the market and traded according to EU regulations and which carry the CE marking. Every mandated product standard includes an annex ZA, which defines the requirements regarding the CE marking and the procedure for the attestation of conformity [29–31]. Attestation of conformity procedure 2S applies to precast concrete products: initial type-testing of the product, factory production control and certification by an approved body (see Fig. 1.4). Construction Products List C contains those construction products that have only a minor significance (e.g. gutters, screeds, etc.). They may not carry the “ mark”, the German symbol of conformity. Annex ZA permits the use of a simplified label for the CE marking according to Fig. 1.5. The details of the product must be stated in an accompanying document according to Fig. 1.6. The design documents mentioned in this are the drawing of the element and the structural calculations. At the time of preparing this chapter (late 2007), only two precast reinforced concrete components may be marketed with the CE marking: – Prefabricated reinforced components made from lightweight aggregate concrete according to DIN EN 1520 – Prefabricated reinforced and prestressed concrete hollow-core slabs according to DIN EN 1168 Consequently, only these currently appear in Construction Products List B Part 1 (edition 2007/1), in section 1.1.6. In addition, the German building authorities now require NABau (Building & Civil Engineering Standards Committee) to draw up a so-called National Application Document (NAD, DIN 20000 -XXX) for every harmonised standard so that the respective Euronorm Table 1.1

Application of the product standard for “precast concrete floors”

Level

General regulations

Product standard Design standard

Europe

DIN EN 13369 Common rules for precast concrete products

Germany

DIN V 20 000-120 Application of building products in structures – Part 120: Application rules for DIN EN 13369

DIN EN 13747 Precast concrete products – Floor plates for floor systems DIN V 20 000-126 Application of building products in structures – Part 126: Application rules for DIN EN 13747

Concrete

EN 1991-1-1 EN 206-1 Eurocode 2

DIN 1045-1

Reinforcement EN 10080 Steel for the reinforcement of concrete

DIN 1045-2 DIN 488 Reinforcing steels/ National Technical Approvals for lattice beams

12

1 General

Fig. 1.5 label

Example of a simplified

Fig. 1.6 Example of the associated accompanying document, see Fig. 1.5

can be used with and is compatible with building regulations in Germany. When an EN standard is introduced, a period of co-existence is defined during which both the DIN standard and the EN standard may be used. A precast concrete floor, and its allocation to national and international design and materials standards, is given here as an example of the practical application of a product standard (see Table 1.1). The period of co-existence for this standard ended on 1 May 2008. Its incorporation in Construction Products List B is imminent [32]. On the national level, the work of the CEN is accompanied by so-called DIN mirror committees, which in the main supervise the work of a TC. There are currently about 80 CEN/

13


TCs active for the construction sector. Those CEN/TCs currently active and relevant to precast concrete construction are listed below together with the standards for which they are responsible. CEN/TC 250 is working on the design standards (Eurocodes), with SC 2 playing the leading role for concrete construction. Since late 2007 all Eurocodes have been available in trilingual editions. Ref. [26] reports on the current status of European standards for concrete, and ref. [27] on the situation regarding reinforcing and prestressing steels. According to the current state of knowledge (late 2008), it will first be possible to use EC 2 with the corresponding German application rules in 2010. CEN/BTS 1 Technical Sector Board for Construction CENTCs (Technical Committees) and the EN standards for which they are responsible and which are relevant to precast concrete construction Standar- TC Object disation No. body

Subcom. Working Gp. Task Group

Status

CEN TC

WG1

TG1

DIN EN 1168

09

TG2 TG3 TG4

DIN EN 12794 DIN EN 12843 DIN EN 13747

07 04 09

TG5

DIN EN 13224

07

TG6 TG7

DIN EN 13225

06

TG8 TG1 TG2 TG3 TG4 TG5 TG6

DIN EN DIN EN DIN EN DIN EN

14992 14843 12737 12893

07 07 07 01

TG9 TG10 TG11 TG12

DIN EN DIN EN DIN EN DIN EN

14258 09 13693 09 14844 13978-1 05

TG13 TG14 TG15 TG1 TG2

DIN EN 14991 DIN EN 15050 DIN EN

07 07

DIN EN

99

229 Precast concrete components relevant to EC 2

Products only partly relevant to EC 2

WG2

Other concrete WG3 products

EN No. Year

1169

Subject/designation

Hollow-core slabs, parts 1 & 2 (CE marking!) Foundation piles Masts and poles Floor plates for floor systems Ribbed floor elements, amend. 05 Ribbed slabs Linear structural elements Wall elements Stairs Floor slats for livestock Elements for fences Vehicle crash barriers Noise barriers Concrete window frames Retaining wall elements Special roof elements Box culverts Precast concrete garages, part 1 Foundation elements Bridge elements Silos General rules for the production control of glass fibre-reinforced cement

14

1 General

Standar- TC Object disation No. body

Subcom. Working Gp. Task Group

Status

DIN EN

CEN TC

EN-No. Year

1170

98–09 Test method for glass fibre-reinforced cement, parts 1-8 07 Common rules for precast concrete products

Framework WG4 DIN EN 13369 guidelines 250 Eurocodes for structural engineering EC1 Safety DIN EN 1990

EC2

SC 1 Actions

DIN EN

1991

SC 2 Concrete construction

DIN EN

1992

02

Basis of structural design 02–95 Actions on structures, parts 1-4 Design of concrete structures

part 1-1 05 part 1-2 09 part 2 04 part 3 07

EC3

EC8

SC 3 Steel construction

ENV

SC 8 Earthquakes

ENV

Subject/designation

1993

General rules and rules for buildings Structural fire design Concrete bridges Liquid retaining and containment structures

part 1-1 05

Design of steel structures part 1-2 05 General rules and rules for buildings Structural fire design 1998 04–05 Design of structures for earthquake resistance

ECISS European Committee for Iron and Steel Standardisation Standardisation body

TC- Object No.

ECISS TC 10

ECISS TC 19

Structural steels

Subcom. Status Working Gp. Task Group TC1

Concrete re- TC1 inforcing and prestressing steels TC2

EN-No. Year

DIN EN 10025

09

DIN EN 10210

06

DIN EN 10219

06

DIN EN 10080

05

DIN EN 10088

05

DIN EN 10138

00

Subject/designation

Hot-rolled products of structural steels, parts 1-6 Hot-finished structural hollow sections of non-alloy and fine-grained steels, parts 1 & 2 Cold-formed welded structural hollow sections of nonalloy and fine-grained steels, parts 1 & 2 Steel for the reinforcement of concrete, parts 1-6 Stainless steels, parts 1-3 Prestressing steels, parts 1&2

2

Design of precast concrete structures

Designing a building made from industrially prefabricated parts calls for certain principles to be adhered to during the planning work (see also [33, 34]). It is important to be familiar with the particular features of precast concrete elements that result from their method of production. Modular dimensions should be defined for the structure and the interior fitting-out and the building divided up into horizontal and vertical grids [35]. The transport dimensions and the loads to be lifted in the factory and on the building site are critical factors for precast concrete buildings. The fire protection, thermal performance and sound insulation requirements as well as the imposed loads for the structural design are all governed by the use of the building. The horizontal stability of multi-storey buildings calls for early coordination, not only with the structural engineer, but with the manufacturer as well. The provision of stiffening cores or walls made from precast concrete components or in situ concrete has far-reaching consequences for the design sequence and on-site construction times. It is advisable to use standardised elements for the loadbearing structure, especially for smaller buildings. Large construction projects tend to generate their own rules and also permit the use of their own systems, although it is then very important to consider the production engineering requirements if an economic design is to be realised. The design of the interfaces between the individual elements is influenced by the structural requirements, but also by the routing of building services. Sensible use of the standard openings provided in the structural carcass or appropriate beam notches is usually only achieved when the structural carcass and the interior fitting-out can be built by the same contractor, i.e. in a turnkey project [10]. The design of the façade determines both the form and the architecture of the building as a whole. In addition, the façade, the “external skin” of the building, must satisfy all the building physics requirements that the environment places upon it. One key design decision is: To what extent does the façade contribute to the loadbearing function? Or is it only a curtain wall? In this book, all these points can only be dealt with in outline. Early planning and coordination of all those involved in the construction project are crucial in order to achieve an optimum building design in terms of architecture, functionality and economics. This begins with the architect and includes the building services and building physics consultants, structural engineers and designers plus the fabrication and erection personnel.


16

2 Design of precast concrete structures

2.1

Boundary conditions for precast concrete design

2.1.1

Production process

The production process for precast concrete components is in many ways fundamentally different to the production process on the building site. For example, columns are mostly cast in horizontal moulds, which means that one side of the column is exposed to the air. If all sides are to have a fair-face finish, then this fourth side requires additional work. Where the column has corbels facing in different directions, then coordination with the factory is required to establish from which side the column can or should be concreted. Walls are mostly cast horizontally on tilt-up moulds, which means that one side is in contact with the mould, the other side is floated. Only in the case of walls produced vertically in battery moulds are both sides in contact with the mould. Façades are generally produced horizontally in a so-called negative mould, i.e. the façade surface is on the underside in contact with the mould. Using this method it is easy to produce textured and exposed-aggregate finishes. Please refer to section 2.4 for the production of sandwich panels (façade panels with integral thermal insulation). As the side panels of moulds on the ground are moved or tilted clear upon demoulding, this joint must be properly sealed for concreting. This is generally achieved with triangular plastic battens, which means that the bottom edges (“bottom” in the sense of the production process) of precast elements are chamfered. There must be a clear indication on the drawings if the top edges are to be chamfered as well. But in many cases beams or T-beams are produced in rigid moulds. In these cases the sides of “rectangular beams” or the webs of double-T sections are slightly inclined outwards so that such elements can be lifted out of the mould once they have hardened without having to move the side panels. This is generally unimportant where the elements will later be concealed behind a suspended ceiling, but where the elements remain exposed, these production-related characteristics of precast concrete elements will need to be considered at the design stage. 2.1.2

Tolerances

The production process gives rise to dimensional deviations of the actual size from the nominal size [36, 37]. For example, dimensional deviations in precast concrete components ensue due to inaccurate transfer of the design dimensions to the mould, deformation of the mould during concreting, deterioration of or wear-related flaws in the mould. However, the production process for a building also includes the erection work, which results in additional positioning tolerances that essentially depend on the methods of measurement employed. In addition, dimensional deviations occur as a result of the deformations of the individual components or the entire structure. These deformations may be load- or time-related (e.g. as result of shrinkage and creep).

17

2.1 Boundary conditions for precast concrete design

DIN 18202 “Tolerances in building construction – Structures” specifies permissible tolerances that apply to the structural carcass and the interior fitting-out irrespective of the building materials. The permissible limits of size for building materials are specified in the materials standards, e.g. DIN 18203-1 “Tolerances in building construction – Part 1: Prefabricated components made of concrete, reinforced concrete and prestressed concrete”, and these must be taken into account as well. According to these standards, there are no longer any accuracy classes as in the past. It has been recognised that the only reason for specifying tolerances in standards should be to guarantee the proper assembly and functionality of components in the structural carcass and interior fitting-out without reworking, i.e. their fitness for purpose, and not, for example, aesthetic demands, e.g. the exact alignment of external wall joints. Fitness for purpose means satisfying, for example, the loadbearing function in the case of short bearing lengths for floor units, or the waterproofing function of an external wall joint. The tolerances laid down in the standards represent the accuracy achievable within the scope of normal diligence. Where greater accuracy is required, then this must be included in the specification, possibly together with the necessary methods of testing and inspection. Greater accuracy causes disproportionately higher costs (see [39, 40] and Fig. 2.1). Tolerances specified in standards should only be understood as production tolerances due to manufacture and assembly. The load- and time-related deformations, just like the fitness-for-purpose requirements (e.g. limit values for the permissible movement of a joint seal), must be limited in other specifications or related to the particular building and taken into account in the structural calculations if necessary. Otherwise the tolerances would only apply for very specific boundary conditions such as the time and date of handover with defined temperature and loading conditions. The tolerance range is the difference between the maximum and minimum sizes. Permissible dimensional deviations of e10 mm therefore translate to a tolerance range of 20 mm (Fig. 2.2). For example, DIN 18202 Table 2.1 specifies permissible limits of size for buildings on plan and elevation (e.g. lengths, widths, grid and storey dimensions) generally applicable to buildings and somewhat higher values for clear opening dimen-

Fig. 2.1 Costs of horizontal building tolerances [39]

18


Table 2.1 Tolerances for prefabricated components of plain, reinforced and prestressed concrete according to DIN 18203-1 a) Limits of size for lengths and widths Line

Component

Limits of size in mm for nominal size in m J 1.5

i 1.5 J3

i3 J6

i6 J 10

i 10 J 15

i 15 J 22

i 22 J 30

i 30

1

Lengths of linear-type e6 components (e.g. columns, beams)

e8

e10

e12

e14

e16

e18

e20

2

Lengths and widths of e8 floor slabs and wall panels

e8

e10

e12

e16

e20

e20

e20

3

Lengths of pre– stressed components

–

–

e16

e16

e20

e25

e30

4

Lengths and widths of e5 façade panels

e6

e8

e10

–

–

–

–

b) Limits of size for cross-sections Line

Component

Limits of size in mm for nominal size in m J 0.15

i 0.15 J 0.3

i 0.3 J 0.6

i 0.6 J 1.0

i 1.0 J 1.5

i 1.5

1

Thicknesses of floor slabs

e6

e8

e10

–

–

–

2

Thicknesses of wall and façade panels

e5

e6

e8

–

–

–

3

Cross-sectional e6 dimensions of lineartype components (e.g. columns, beams, ribs)

e6

e8

e12

e16

e20

c) Angular tolerances Line

Component

Angular tolerances as perpendicular measurements in mm for length L in m J 0.4

i 0.4 J 1.0

i 1.0 J 1.5

i 1.5 J 3.0

i 3.0 J 6.0

i 6.0

1

Wall panels without finished surface and floor slabs

8

8

8

8

10

12

2

Wall and façade pa5 nels with finished surface

5

5

6

8

10

3

Cross-sections of lin- 4 ear-type components (e.g. columns, beams, ribs)

6

8

–

–

–


Fig. 2.2

19

Terminology of tolerances

sions (e.g. between columns) plus limits of size for window or door openings depending on the nominal sizes. Limit values are also specified for angular and flatness deviations and also out-of-plumb deviations for columns, checked by way of permissible perpendicular measurements (DIN 18202; see Tables 2, 3 and 4). These may no longer be added to the limits of size. This corresponds to the box principle of ISO 4464 (now withdrawn), according to which the actual dimensions of a component or an opening must always lie within the limit dimensions (Fig. 2.3). The permissible flatness deviations do not include the flatness of the components with respect to each other, which must be considered additionally. For example, the steps be-

Fig. 2.3 Illustration of the box principle using the example of permissible dimensional deviations of openings (permitted deviations and angular tolerances) [37]

20


tween adjacent prestressed concrete planks are often unavoidable and the admissibility of such steps must be regulated separately. By contrast, DIN 18203-1 (see Table 2.1) specifies the manufacturing tolerances for the precast concrete components themselves, divided into limits for length, width and cross-sectional sizes of linear elements or floor, wall and façade panels. The angular tolerances for planar panels and slabs and the cross-sections of linear components are also shown. Ref. [36] is a commentary on DIN 18201 and DIN 18202. It contains advice for planners concerning tolerances and also proposes a method for checking the alignment of columns in frame structures and single-storey sheds. Structures with accuracy requirements according to DIN 18202 should always be checked and monitored using surveying techniques. Conventional measurements by the foreman using a profile board, line and extending tape measure are by no means adequate! However, the German standards do not include any details about permissible deviations for measurements. pffiffiffi According to ISO/DIS 4463, limits of size of e2K · L [mm] (distance L in m) at intervals of i 4 m are permissible (see also [37]), where K w 5 for earthworks, and K w 2 for structural works. For many situations, defining minimum requirements for tolerances according to the standard is adequate in practice. However, this does not necessarily mean that this is adequate for the “fitting together”. That can only be established following an appropriate fit calculation which, however, presumes knowledge of the production accuracy achievable. And the tolerances specified form the foundation for this. Whether the manufacturer of the precast concrete components also erects and assembles these is also a crucial factor. Where this is not the case, all subcontractors would insist on the inaccuracies to which they are entitled and only the additive method remains if disputes are to be avoided. Fit calculations taking into account the law of the propagation of uncertainties can certainly yield savings, e.g. in jointing materials, for contractors who have the entire production process (measuring, production and erection of the precast concrete components) under control in terms of tolerances. Examples of such calculations can be found in [37, 41, 42]. Furthermore, with structures built from precast concrete elements, the tolerances at the supports are especially important. It must be guaranteed that the as-built tolerances match those on which the structural calculations were based. Permissible tolerances that influence stability must therefore be specified on the working drawings. The tolerances of built-in items and connectors are specified in [38] together with a simple method for a fit calculation.

21


2.1.3

Transport and erection

Dividing a structure into prefabricated elements is to a large extent governed by the transport restrictions and the erection weights of the individual components. The aim should be to make the elements as large as possible because every subdivision doubles the handling activities necessary in the factory and on the building site. The higher the quality of an element, i.e. the more fitting-out components it contains (e.g. windows, doors or building services in a wall panel) or the more functions it has to perform (e.g. a façade element providing loadbearing, thermal insulation and architectural functions), the lower is the percentage cost of the transport. It is the permissible road transport dimensions according to Germany’s Road Traffic Act (StVZO, Straßenverkehrs-Zulassungs-Ordnung) that have led to today’s typical widths for floor units of 2.40 or 2.50 m and wall panel heights of I 3.60 m [43]. Where dimensions or total weights exceed those given in Table 2.2, then a special permit according to StVZO cl. 29 must be applied for, and a police escort may even be necessary. Such permits can be issued by the respective authorities (e.g. local government departments) for each individual case or as a general permit valid for several years. Individual permits will be necessary where the dimensions of the precast concrete components exceed the dimensions given in Table 2.2. In such situations it is essential to establish the potential transport route at an early stage and also the duration and time of the delivery (maybe only during the night). And when an oversize load has to travel through more than one federal state, then a travel permit must be applied for in each state and the various permits coordinated. This can prove to be extremely complicated in some cases, with negative repercussions for costs and delivery times. The vehicle types given in Table 2.3 are generally used for road transport. Transport by rail is relatively rare – apart from projects for the railways themselves – because the transfer from road to rail and back again before the building site is reached is usually unavoidable. And the prerequisite in every case is that the factory itself has a direct railway link. Transport in containers, in which width and height are limited to approx. 2.30 m and the length to 12.00 m, are hardly relevant for structural precast concrete elements. The reader should refer to [48] for information on the problems of transport across national and international frontiers. Table 2.2 Maximum permissible dimensions and total weights for road transport (depends on particular approving authority) Without special permit (to StVZO cl. 32)

With annual permit (StVZO cl. 29)

Width

2.55 m

3.00 m

Height

4.00 m

4.00 m

Length

15.50 m

24.00 m

Total weight

40 t

48 t (tractor unit with self-steering trailer)

22 Table 2.3


Vehicle types for road transport

Type of component

Type of transport

Columns and beams I 16 m long

Tractor unit with (extending) semi-trailer

Columns and beams j 16 m long

Tractor unit with trailing bogie

Façade panels

Low-loader with frame for panels

Ground floor panels and ground beams

Tractor unit with (low-bed) trailer

Bridge beams

Tractor unit with trailing bogie

It is also vital to consider every detail of the erection sequence when designing the elements. The type of erection must be taken into account: horizontal, i.e. elements positioned storey by storey with a tower crane, or vertical, i.e. bay by bay over the full height of the building with a mobile crane (Fig. 2.4). Typically, tower cranes can handle only relatively light loads, albeit at a large radius and through a full 360h. However, the largest tower crane used in Germany to date was able to handle a load of 30 t at a radius of 40 m. Mobile cranes can lift heavy elements, but only from a position with a firm, stable base. And owing to their limited working radius and restrictions on the slewing circle with outriggers extended, they often have to be repositioned during the work. These days, mobile cranes with a capacity of 400 t are relatively inexpensive. This is because it is the hire period and not the actual cost of the crane itself that governs. For example, it takes almost a whole day to reposition a 500 t crane. Cranes mounted on crawler tracks are used where even greater lifting capacities are required. Although crawler-mounted cranes with lifting capacities of up to 1300 t take approx. 1–2 weeks to set up, their crawler tracks mean that

Fig. 2.4

Types of erection and typical crane dimensions with loads


23

precast concrete elements are easily transported and positioned anywhere on the building site, provided adequate manoeuvring space is available. Of course, both types of erection can be combined on one building project, and adapted to suit the conditions, with the tower crane remaining on site for the duration of the entire works and mobile cranes being hired by the day as required. One interesting example of such a detailed coordination of both forms of erection can be seen in Fig. 2.5, Zu¨blin House. On this project, erection was divided into four phases (see also Figs 2.125 and 2.149) [44]. Precast concrete components are being increasingly incorporated into composite precast/ in situ concrete designs. In this approach it is possible to exploit the advantages of prefabricated production (complex geometry, surface finish, formwork savings for large series, etc.) and build the precast concrete elements into an in situ concrete structure. Attention should be given to making sure that the elements are not too heavy to be positioned with a tower crane. If this is not possible, then the use of an additional mobile crane should be concentrated into one period because otherwise the cost of having two cranes on site will make itself felt. 2.1.4

Fire protection

Besides ensuring adequate stability, durability, thermal performance, moisture control measures and sound insulation, it is also vital to verify the fire resistance, especially for the loadbearing and enclosing components. This is carried out with DIN 4102 “Fire behaviour of building materials and building components”, which is discussed in detail in [45]. The design rules are based on an internationally agreed standard temperature curve used in many countries. DIN 4102-1 allocates building materials to classes according to their reaction to fire (Table 2.4). Building materials class A1 is for those materials that are incombustible in the classical sense, e.g. concrete and steel. Class A2 covers newer building materials that contain combustible constituents to some extent, e.g. the majority of gypsum-based boards or polymer concretes.

Table 2.4

Building materials classes to DIN 4102-1

Building materials class

Building authority designation

A

Incombustible building materials

B

A1 A2 B1 B2 B3

Combustible building materials Not readily flammable building materials Flammable building materials Highly flammable building materials

24


(a)

(b)

(1a)

(1b)

Fig. 2.5 An erection sequence divided into four phases using the example of Zu¨blin House. Phase 1: Vertical erection of columns with mobile crane; (a) section through building, (b) positions of tower cranes and slewing circles for horizontal erection. (1a) Erection of façade columns, (1b) erection of internal columns.


(2a)

(2b)

(2c)

(2d)

25

Fig. 2.5 Phase 2: Horizontal erection of perimeter beams, inverted channel section floor units and floor planks with four tower cranes. (2a) Erection of L-shaped perimeter beams on façade columns, (2b) erection of inverted channel section floor units on internal columns, (2c) positioning the floor planks, (2d) concreting the floor slabs.

Fig. 2.5 Phase 3: Vertical erection of curtain wall façade bay by bay.

Fig. 2.5 Phase 4: Vertical erection of underground car park, atrium lift/stairs tower, atrium walkways and roof frame with two heavy-duty telescopic cranes and one tower crane.

26


One typical example of a not readily flammable building material (class B1) is the lightweight wood-wool board. Certain testing regulations for furnace tests apply for the classification of the building materials. Joint sealing compounds or strips belong to class B1 or B2 depending on their composition. They may be incorporated between concrete components in certain minimum depths and maximum joint widths. Elastomeric bearings fall into class B2. Only class A1 materials may be used for the sealing materials in expansion joints that must satisfy fire protection requirements, e.g. mineral-fibre boards, asbestos foams or fibres and aluminium silicate fibres (see Fig. 2.13). Components are classified according to their fire resistance; the fire resistance ratings are given in Table 2.5. The fire resistance of components is therefore specified according to fire resistance rating and building materials class. For example, the abbreviated form for a fire resistance of 90 minutes is F 90. Suffixes A, B, or AB are added to designate the combustibility: F 90 -B:

general

F 90 -AB: essential components incombustible (loadbearing structure and enclosing elements) F 90 -A:

all components incombustible

The current regulations for multi-storey buildings generally specify an F 90 rating for loadbearing components. The commonest requirements for building components are F 30 -A and F 90 -A. The loadbearing structure of a high-rise building must comply with F 120 -A above a height of 60 m. And even F 180 -A above a height of 200 m (see also the High-Rise Building Directive of the Hessen Ministry of the Interior). DIN 4102-3 contains further requirements (e.g. additional impact loads) for fire walls and non-loadbearing external walls, which include spandrel and fascia panels as well as room-high, room-enclosing external walls. Parts 5 to 8 of DIN 4102 – mentioned here for completeness although they apply less to concrete construction and more to building services and interior fitting-out – deal with the fire protection of fire stops, lift enclosures, glazing and ventilation ducts and assign them appropriate fire resistance ratings (e.g. T 90, G 90, L 90, K 90, where T w door, G w glass, L w ventilation and K w flaps). The resistance of roof coverings to flying sparks is another aspect covered by these parts. In Germany the fire protection requirements are generally defined in the building regulations of each federal state together with the provisions for their implementation. However, the terms fire-retardant, fire-resistant, etc. are used here, which must be allocated to the respective DIN 4102 terms in the introductory decrees, as listed in Table 2.5. The federal state building regulations only cover standard buildings for standard uses (e.g. housing and offices) and therefore special facilities for special uses are dealt with in special legislation. The following are just a few examples:

27


Table 2.5

Fire resistance rating F and building authority designations

Fire resistance rating to DIN 4102-2

Duration of fire resistance in minutes

Building authority designation according to introductory decree

F 30 F 60 F 90 F 120 F 180

i 30 i 60 i 90 i 120 i 180

fire-retardant fire-resistant highly fire-resistant

– Gescha¨ftshaüserverordnung (GhVO, Business Premises Act), which covers, for instance, department stores, supermarkets, etc. – Versammlungssta¨ttenverordnung (VSta¨tt-VO, Places of Assembly Act), which covers, for instance, lecture theatres, sports halls, etc. – Garagenverordnung (GarVO, Garages Act), which covers, for instance, small garages, multi-storey car parks, etc. – Schulhaus-Richtlinien (SHR, School Buildings Directive) – Industriebaurichtlinie (IndBauR, Industrial Buildings Directive) The last of these refers to DIN 18230 “Structural fire protection in industrial buildings”. Part 1 of this standard contains a method of calculation that allows industrial buildings with definable fire loads to be designed with respect to the theoretically necessary fire resistance of their components if required – a different approach from that in the Industrial Buildings Directive. As precast concrete components by their very nature provide a high level of fire resistance, such verification is generally unnecessary. Further information on fire protection for industrial buildings can be found in [46]. There are also special tall building and school building directives which, however, are not legally binding in all Germany’s federal states. Where reinforced concrete components are subjected to compression, it is the concrete’s reaction to fire that is particularly relevant [49]. But where components are subjected to bending or tension, it is primarily the strength and deformation behaviour of the reinforcing steel. According to DIN 4102- 4, the critical steel temperature critT is the temperature at which the yield strength of the steel drops below the steel stress in the component; critT w 500 hC for reinforcing steel, and all the design rules are based on this. For prestressing steels (e.g. cold-drawn strands with critT w 350 hC), please refer to DIN 4102- 4 Table 1 (see also [47]). The compressive strength of the concrete is also dependent on the temperature: it drops to approx. 70 % at 200 hC, and at 750 hC is only about 20 % of its strength at 20 hC. However, knowledge of the temperature distribution within the cross-section is important for reinforced concrete components because the edge distances for the reinforcement are based on this (Fig. 2.6 shows an example).

28


Fig. 2.6 Isotherms in hC for a T-beam exposed to fire [45]

Section 2.6.5 deals in more detail with the design of individual precast concrete elements to meet fire protection requirements. 2.2

Stability of precast concrete structures

The fundamental considerations regarding the stability of frame building structures are described in detail in [50]. A number of general thoughts on this subject are summarised below and the problems specific to precast concrete construction examined in more detail. 2.2.1

Arrangement of stability elements

Stability in residential and office buildings is generally assured by stair shafts and/or enclosing shear walls. By contrast, in precast concrete single-storey sheds intended to house production processes and some precast concrete frame structures with one or two storeys, the horizontal stability is provided by the columns. The columns of such buildings usually extend over the full height of the building and are fixed at their foundations; the beam-column connections in such buildings are pinned. Such systems are classed as sway, or unbraced, frames and must be designed according to second-order theory taking into account the deformed system (Fig. 2.7). Structures with more than two storeys require additional shear walls, frames, girders or torsion-resistant service cores to ensure their horizontal stability. The connection of series of pinned-end beams and columns to the stability components is achieved via the relatively rigid floor diaphragm.

Fig. 2.7

Sway systems (design according to second-order theory)

2.2 Stability of precast concrete structures

Fig. 2.8

29

Arrangement of building stability elements on plan

When planning stabilising shear walls or cores, the aim should be to create a statically determinate arrangement on plan in order to prevent restraint forces in the floor diaphragms as a result of shrinkage or temperature changes. In addition, it is essential to ensure that the stabilising cores or shear walls are positioned in a way that minimises the rotation of the building on plan in the case of a uniformly distributed horizontal load due to wind and eccentricity. Shear walls must be positioned in at least two non-parallel directions and on at least three axes (Fig. 2.8). With statically determinate bracing systems, it is the deformation capacity of the columns that governs the maximum building size for a frame structure. According to [50], lengths of 100 m and more are possible without expansion joints in a frame structure. The deformation capacity of the columns depends on the accuracy of the representation of the stiffnesses in the cracked condition. Critical here are the cross-sectional values but, first and foremost, the magnitudes of the axial forces to be carried by the columns [52]. The deformation capacity of the columns can be increased by including pinned supports for the columns or sliding bearings for the first floor slab above the underside of the foundation, although such measures only need to be provided at the columns furthest from the core (Fig. 2.9). In the case of statically indeterminate bracing systems, non-uniform temperature changes result in restraint forces between suspended floors and stability components (Fig. 2.10) (see section 2.2.2.5). Joints are usually the most appropriate means of avoiding restraint forces provided simple and clearly arranged joints are possible (Fig. 2.11). Fig. 2.12

30


Fig. 2.9 Measures for increasing the deformability of columns and walls for horizontal floor expansion

Fig. 2.10 Restraints caused by the prevention of horizontal floor deformation

shows possible joint arrangements, and Fig. 2.13 shows joints that satisfy fire protection requirements, where such are necessary. Expansion joints always represent a source of potential damage – details that require substantial input to rectify. Their design should therefore be carefully thought out and the appropriate drawings and installation instructions must be prepared. Quality control on the building site is essential for guaranteeing proper construction. Damage often occurs as a result of unintentionally filling the entire joint with concrete (constructional measures to prevent loss of grout) and incorrect installation or the installation of the wrong bearings.

31


Fig. 2.11

The principles for positioning joints

Fig. 2.12

Fig. 2.13 Joints complying with fire protection requirements [45]

Building joints

32


Therefore, constructing buildings without joints is something that should always be considered. Where a building is to be constructed without joints, then the following aspects in particular must be considered: – The actual expansion due to temperature changes and shrinkage – The deformability of the stability components (including horizontal precast concrete floors), especially in the cracked state – The creep deformability of the concrete – The conditions during construction A careful study of the problem using modern methods of calculation often leads to expansion joints being omitted. At the very least a check should be carried out to ascertain whether the joints need to be continued over the full height of the building or whether the upper floors of tall buildings could be built without joints (Fig. 2.14) [53]. It is always only the lowest floors that are affected most by restraint forces (if we neglect the fire loading case in the upper storeys). In Zu¨blin House (Fig. 2.15) each of the two 94 m long blocks is divided by one joint. This positioned the cores somewhat off centre and so these were strengthened by two crosswalls on the two central axes. The two floor diaphragms were connected together near one of these two walls by means of a shear key (joggle or castellated) joint that is able to move in the longitudinal direction of the building, which means that both floor diaphragms can be supported on the cross-walls. The topmost floor of the building was built without any joints. The ensuing restraint forces can be accommodated by this floor slab and the cores. Stiffening shear walls can also be offset storey by storey, although in that case the shear forces in the walls will have to be transferred via the corresponding floor diaphragms (Fig. 2.16). The forces, and the floor deformations in particular, must be taken into account in the structural calculations. It must be guaranteed that the forces can be transferred between the vertical and horizontal stability elements. Openings in the floor are common at shear walls and service cores in particular, and these hinder the transfer of forces. The shear wall moment cannot be carried via the floor diaphragm acting as a mem-

Fig. 2.14 only

Expansion joint positioned in the lower storeys

33


Fig. 2.15 Building stability and joint positions for Zu¨blin House [44]

① Stiffening walls ② Building joint with shear key ③ Horizontal forces from atrium roof beams in floor above 5th storey ④ Post-tensioned prestressed concrete for walkways above 5th storey ⑤ Service cores for stability

Fig. 2.16

Offset positioning of shear walls

brane, instead must be transferred to the foundations via the neighbouring columns in the form of a couple. Shear walls offset in different storeys must therefore match the structural grid. 2.2.2

Loads on stability elements

2.2.2.1

Vertical loads

The vertical loading due to dead and imposed loads is carried by the columns, walls and service cores. The cores and stiffening walls should carry permanent vertical loads wherever possible in order to do justice to their function (Fig. 2.17). Asymmetrical load transfers result in vertical loads acting eccentrically, which leads to moments being applied at the underside of the foundation. Eccentric loads on columns can also lead to horizontal loads being exerted on the core as a result of tie forces (Fig. 2.18). These can be avoided by cantilevering the beams beyond the columns, as shown in Fig. 2.19.

34


Fig. 2.17 Core wall, subjected to vertical and horizontal loads

Fig. 2.18 Horizontal loads on core due to restraining forces caused by eccentric column loads

Fig. 2.19

2.2.2.2

Centring eccentric column loads

Wind loads

Wind loads are determined according to DIN 1055- 4. In contrast to previous editions of DIN 1055, the provisions now also apply to buildings susceptible to vibration up to a height of 300 m. Almost all engineered structures (with the exception of bridges, but including chimneys) are now covered as well. The classification of wind speeds on the European wind map also guarantees pan-European continuity with respect to the loading assumptions. Besides improved aerodynamic coefficients, there is now a distinction between inland and coastal regions, and the influences of terrain roughness and turbulence-induced transverse vibrations are also dealt with. Cyclic wind loads can induce vibrations in structures. Such vibrations lead to an increase in the loads due to wind pressure or suction. The susceptibility to vibration does not need to be considered when the increase in a deformation due to gust resonance does not exceed 10 %. DIN 1055- 4 specifies simplified determination criteria for this. As a rule, residential, office and industrial buildings up to 25 m high, and other buildings similar in form and type of construction, can be assumed to be not susceptible to vibration, and separate verification is unnecessary. However, buildings whose stability relies on fixedbase columns are relatively flexible and their susceptibility to vibration should be checked even for building heights I 25 m. The main calculation steps for these buildings are summarised below. The special provisions of the standard must be taken into account for buildings with a special form, taller buildings or buildings in exposed locations (e.g. coastal sites).

35


Fig. 2.20

Pressure coefficients to DIN 1055-4

Fig. 2.21 Interpolation of pressure coefficients depending on reference area to DIN 1055-4

With an orthogonal arrangement of the stability elements, i.e. the normal case, the wind loads are examined separately for the two main axes of a building. These add up to the following for the entire structure: FW w cf qðze Þ Aref

(1)

where cf ze Aref q

aerodynamic force coefficient reference height reference area for force coefficient dynamic pressure

The pressure coefficient cpe, which depends on the reference area, should be used for cf. This figure can be read off from the table in Fig. 2.20, which depends on the area of the building on the windward side and the ratio of building height to building depth (h/d). Intermediate figures between 1 and 10 m2 can be obtained by linear interpolation (Fig. 2.21). The reader is referred to DIN 1055- 4 for suction loads perpendicular to the direction of the wind. In the normal case and when assuming a mixed terrain profile in categories II and III (mixture of individual structures and suburban development, also industrial areas), the following pressure profile can be assumed for the velocity-related wind pressure q(ze):

36


Table 2.6 Simplified dynamic pressure assumptions for buildings up to 25 m high to DIN 1055-4 (only wind zones 1 & 2 shown) Wind zone

1 2

Dynamic pressure q in kN/m2 for building height h h J 10 m

10 m I h J 18 m

18 m I h J 25 m

Inland

0.50

0.65

0.75

Inland

0.65

0.80

0.90

Baltic Sea coast and islands

0.85

1.00

1.10

qðze Þ w 1,5 qref for z J 7 m z 0,37 for 7 m I z J 50 m qðze Þ w 1,7 qref 10 The reference pressure qref is 0.32 kN/m2 for wind load zone 1 and 0.39 kN/m2 for zone 2. Wind load zones 3 and 4 must be considered for sites near the coast. For locations i 800 m above sea level, the wind pressure should be increased by 10 % for every 100 m rise in altitude. For simplicity, the wind pressure may also be assumed to be constant for buildings J 25 m high. The values given in Table 2.6, which is based on Table 2 of DIN 1055- 4, then apply. The example of a 20 m high building in wind load zone 2 shown in Fig. 2.22 indicates that the simplified assumption of a constant wind pressure results in higher wind loads. Generally, but especially for buildings with asymmetrically arranged shear walls, it should be noted that the wind load must be applied eccentrically, with an eccentricity of: ew

b 10

This can lead to appreciable torsion loads in the case of stiffening service cores.

Fig. 2.22

Comparison of simplified dynamic pressure assumption q and standard case

(2)

37


2.2.2.3

Out-of-plumb loads

As a substitute for dimensional deviations of the system during construction and unintentional eccentricities of the load application, DIN 1045-1 specifies that an out-of-plumb deviation for the centroid axes of all columns and walls can be taken into account. This loading case must be calculated with the full load. It is regarded as an independent loading case and must be considered for the ultimate limit state except for accidental actions. This means that loads due to wind or earthquake must be added into this calculation. The effect of the dimensional deviation may be replaced by the effect of equivalent horizontal forces. The out-of-plumb deviation for floor diaphragms may be considered by an angled position aa2 corresponding to Fig. 2.23. This approach has been used in DIN 1045-1 (in turn taken from [54]). In the latter publication it was established that a uniform inclination of the columns becomes less and less likely as the number of columns increases. This was made clear by measurements carried out on precast concrete frame structures. Anyway, the approach according to DIN 1045-1, which assumes storey-high pinned columns, is not a realistic theoretical model for the continuous columns normally used in precast concrete construction. For that, a column system line without kinks would be more reasonable. The resulting forces Hfd are transferred to the floor diaphragms and via these to the shear walls. But their further transfer to the vertical stability components does not need to be verified by calculation. For the design of the vertical components, an inclined position aa1 is to be assumed for all vertical components, i.e. the stabilised and the stabilising components, according to Fig. 2.25. DIN 1045-1 quite rightly takes into account the fact that it is unlikely that construction inaccuracies would continue uncorrected to the top of the building. As already mentioned

Fig. 2.23 Out-of-plumb loading case to DIN 1045-1 (for floor diaphragms)

38


Fig. 2.24

Inclined position and reduction factor as functions of the number of columns

Fig. 2.25

Out-of-plumb loading case to DIN 1045 (for vertical stability components)

above, the decreasing likelihood that all adjacent columns have an identical inclination means that the angle may be reduced by a coefficient an. However, this means a maximum reduction of only a little less than 30 % (Fig. 2.24).


2.2.2.4

39

Seismic loads

Most earthquake damage is caused by seismic activities near the earth’s surface (tectonic earthquakes). The jerky movements in the earth’s crust cause energy to be released in the form of seismic waves. Damage to structures is caused by vibrations transferred from the ground to the structure [55]. Whereas vertical ground movements increase the vertical loads only marginally and therefore can be ignored, horizontal ground accelerations can cause a comparatively large increase in the horizontal loads. These horizontal loads depend on the magnitude of the ground accelerations in the subsoil, the natural frequencies and, primarily, the mass of the building. The publication of DIN 4149 in 2005 adapted the 1981 edition of the same standard to the European design approach. In Germany itself, horizontal loads as a result of earthquakes only need to be considered in a few regions. But the increase in the overseas activities of Germany’s construction industry has added more significance to the issue of protection against earthquakes. Compared to the previous edition of the standard, the new DIN 4149 takes a more differentiated look at the following aspects: – The construction of the structure – Ground and subsoil influences – The significance of the structure itself – Torsional vibrations – The ductility of the structure – Methods of calculation The seismic loads on the structure are therefore now higher than was the case according to the old edition of the standard – in some cases noticeably higher than the horizontal loads due to wind. This in turn results in both favourable and unfavourable aspects for the design of structures made from precast concrete elements. The favourable aspects are the lower masses of precast concrete structures compared to in situ concrete ones and also the lower loads due to a relatively flexible stability system with fixed-base columns. On the other hand, rigid stability systems with shear walls lead to higher horizontal loads. Also unfavourable are the low-ductility connections between the precast concrete components. As a rule, precast concrete structures have only a few energy-dissipating components available, which means that they practically elastic behaviour must be assumed during earthquakes, which in the end leads to relatively high equivalent loads. The procedure for verifying stability under seismic loads does not essentially differ from the procedure used in the past. For simplicity (and this is generally the case), a static equivalent load may be assumed to replace the dynamic vibration processes. This results in the following: Fb w Sd ðTÞ M

(3)

40


Fig. 2.26 Response spectrum for different ductility values

where the static total load Fb and the design acceleration Sd are functions of the natural period of vibration of the structure. The design acceleration (design response spectrum) is obtained from the elastic response spectrum taking into account the non-linear, or rather ductile, behaviour of the structure. The response spectrum here describes the magnitude of the maximum response (e.g. acceleration) of a linear elastic single-mass vibrator, with period of natural vibration T, to the design earthquake. Fig. 2.26 shows the example of a design acceleration plus the influence of the structure’s ductility. The response spectrum for determining the accelerations of the structure is generally the governing factor. It is principally dependent on the following parameters: x

Seismic zone and resulting ground acceleration zone 0 – 0 m/s2 ground acceleration 1 – 0.4 m/s2 2 – 0.6 m/s2 3 – 0.8 m/s2

x

Subsoil and ground classes classes A, B, C and R, S, T

x

Importance category of building category I e.g. agricultural buildings category II residual buildings category III schools, department stores category IV hospitals, safety/security facilities

x

Structure ductility behaviour classes 1 and 2

The corresponding ductility (behaviour) factor is defined as follows: qw

Rel Rnl

(4)


41

Fig. 2.27 Distribution of total seismic load over height of building

and corresponds to the quotient resulting from the elastic component resistance and the non-linear (ductile) resistance. For reinforced concrete structures this value lies between q w 1.5 (e.g. stability provided by walls) and q w 3.0 (e.g. stability provided by frame). DIN 4149 lays down appropriate constructional rules for this. A lower ductility (q w 1.5 or even 1.0) should be assumed for structures built from precast concrete elements. In particular, a value of q J 1.5 should always be assumed for the design of foundation elements. The total seismic load determined in this way can be apportioned according to the eigenmode or, for simplicity, linearly, taking into account the respective storey mass (Fig. 2.27). This simplified approach only applies to regular systems that can be calculated as plane systems in both directions. A three-dimensional calculation is necessary for irregular systems. The torsional vibrations of the building must be considered in addition to the horizontal loads. If the system is almost symmetrical on plan and on elevation and it can be assumed that the centre of gravity of the mass coincides approximately with the centre of gravity of the stability elements (centroid of horizontal loadbearing elements), then an accidental torsion load may be considered in the following simplified form: x (5) d w 1 S 0,6 Le Here, the horizontal load of the individual stability components must be increased by the factor d. Fig. 2.28 illustrates the calculation graphically.

Fig. 2.28 Determination of factor d for taking into account torsional vibration

42

Fig. 2.29


Eigenmode for a building calculated with the help of an FEM analysis

All methods of calculation based on the response spectrum method are permissible. The methods can be subdivided into the simplified response spectrum method, which is based on a single-mass vibrator and must comply with certain application rules, the multi-modal response spectrum method, which is based on a multiple-mass vibrator and takes into account several eigenmodes and mass participations (modal masses), and the three-dimensional modelling of the building, taking into account the true mass distribution and the acceleration values for the individual stability elements. The delayed occurrence of individual maximum loading values can be taken into account in this last method. Fig. 2.29 shows an example of the first eigenmode of a building based on a three-dimensional analysis. The equation for determining the period of natural vibration given in the 1981 edition of DIN 4149 can still be used for the simplified method: s ffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi X n H 1 mi z2i (6) T1 w 1,5 S 3EI Ck JF iw1 Therefore, the influence of foundation rotation can also be taken into account approximately for pad foundations. The dynamic subgrade modulus – which is much higher than the static subgrade modulus – must be used as well as the moment of inertia of the foundation. Foundation rotation does not normally have to be considered for buildings that rely on service cores and shear walls for their stability. Using this period of natural vibration it is possible in the simplest case, with the associated response spectrum and ductility factor, to determine the static equivalent load and apportion it to the storeys according to Fig. 2.27. The (internal and external) stability must be verified for the static equivalent loads calculated. The earthquake loading case combination according to DIN 1055-100 must be applied here. Reduced imposed loads must be considered in addition to the dead loads. The combination coefficient c2 (DIN 1055-100) can be further reduced by the factor f (DIN


43

4149). The snow loading case must also be considered among the variable loads when carrying out the seismic design. The most unfavourable combination is required when considering seismic actions in two planes (x- and y-directions): 1.0 · Ex with 0.30 · Ey or 0.30 · Ex with 1.0 · Ey. This is generally only critical when stability is provided by fixed-base columns. The factor of safety for the actions is gE w 1.0, whereas the factor of safety for the materials should be taken as gM w 1.5 for concrete and gM w 1.15 for reinforcing steel. Where stability is provided by way of fixed-base columns, these should be checked for buckling, also for the earthquake loading case. For simplicity, this analysis is not essential when the horizontal load due to seismic actions is dominant, i.e. when the following condition is satisfied: uw

Ptot df J 0,10 Vtot h

(7)

where Ptot Vtot h df

vertical load associated shear force due to earthquake column height in storey relative horizontal deformation in storey due to earthquake

The latter can be calculated from the horizontal deformations assuming elastic material behaviour, increased by the ductility factor q used earlier. Great attention must be paid to the detailed design of the structure. The following points are especially important for precast concrete elements: – Structural connections at all supports for precast concrete components (e.g. pin shear connectors). – Securing of “non-loadbearing” components. – Floors designed as horizontal diaphragms. – Structural and preferably ductile connections to bracing and stiffening components. – Structural connections between foundation components so that no relative deformations ensue between foundations; exceptions are possible depending on the subsoil (see DIN 4149 section 12.1.2). – Ductile and load-carrying design of core walls with openings, “soft” storeys in particular, which can lead to a storey-by-storey failure of the entire stability element, should be avoided. – Use of highly ductile steel in the tension zones of stability elements; walls and floors may be reinforced with steel of normal ductility, likewise shear links and lattice beams in the floors.

44 2.2.2.5


Restraint loads (shrinkage and temperature)

Shrinkage and temperature changes in the floor diaphragms can cause restraint action effects in vertical loadbearing elements (columns, cores, walls) where the latter prevent unrestricted deformation of the floor diaphragms. According to DIN 1045-1 section 7.1, shrinkage deformations must be taken into account when they are significant for the loadbearing structure. In doing so, it should be noted that shrinkage stresses are considerably dissipated by creep and so only a very much reduced rate of shrinkage can be considered. Furthermore, the rate of shrinkage should be distinguished according to part of building and use of building. In particular, components buried in the ground with a damp climate exhibit smaller shrinkage strains, whereas a higher rate of shrinkage should be assumed in very dry climatic conditions, e.g. on permanently heated retail premises. In addition, in the case of precast concrete structures, by the time the floor diaphragms are grouted in place, a large proportion of the shrinkage has already taken place. Whereas with exclusively precast concrete structures the deformations can be primarily accommodated in the resilient joints, in composite precast/in situ concrete structures it should not be forgotten that because of the different concrete (shrinkage) ages, the semi-finished component restricts the shrinkage of the in situ concrete. Shrinkage cracks are therefore generally found in the joints between the semi-finished components. The disadvantage of this is that the entire shortening collects in single cracks, but the advantage is that the location of the cracks is very likely to be known and appropriate “anti-crack” reinforcement can be provided in the joints between the semi-finished components. When analysing the internal forces or deformations caused by temperature changes, which may be necessary with very long structures, DIN 1055-7 specifies that for buildings the temperature may be assumed to be constant throughout the entire loadbearing structure. For those components protected against temperature changes, as is the case for suspended floors in thermally insulated buildings, average temperature fluctuations of max. e7.5 K may be assumed for simplicity. Special considerations regarding the temperature differences to be assumed are necessary for external components (e.g. parking decks). The coefficient of thermal expansion of normal-weight concrete should be taken as aT w 10 –5 · K–1, that of lightweight concrete aT w 0.8 · 10 –5 · K–1. The restraint action effects caused by shrinkage and temperature fluctuations (the shortening of the components is critical) essentially depend on the stiffness of the stability components and the floor linking these. The greater the resilience of the stability components, the lower are the restoring forces generated. Expansion joints must be introduced if the restraint action effects can no longer be accommodated. In any case, however, it is essential to consider the deformations and restraint effects as realistically as possible. Designers always try to use the stability elements to accommodate the restraint effects and thus avoid expansion joints. Normally, the restraint forces are calculated taking into account the cracked components. In doing so, the highly stressed force transfer zones and the local cracking at these points must be especially considered.


45

Temperature loads are variable action effects according to DIN 1055-100. With an elastic calculation of the restraint forces, the factor of safety according to DIN1045-1 section 5.3.3 may be taken as gQ w 1.0. If the analysis is carried out taking into account the cracked components, a factor of safety of gQ w 1.5 should be used. 2.2.3

Distribution of horizontal loads

The development of powerful computers and the availability of FEM programs mean that these days complete buildings can often be calculated as a whole. All the loadbearing components of the building can therefore be modelled in detail. Openings in shear walls can be taken into account, likewise the distribution of the horizontal loads depending on the stiffnesses of the walls. However, care is required because only a few programs can model the conditions during construction properly. Cracking and the redistribution of loads as a result of creep and shrinkage are generally not taken into account, and restraint action effects due to hydration of the concrete, temperature loads and settlement of the structure are usually totally ignored. A plausibility check of the results often does not even allow extensive combination of individual loading cases to form complex action effect situations. Three-dimensional calculations of complete buildings therefore involve considerable risks for the design of the components if the plausible and traceable derivation of the action effects is not possible, and checking the results is totally impossible. An understanding of the loadbearing behaviour and the experience of the engineer are therefore gradually disappearing. Therefore, the recommendation is to distribute the horizontal loads as described below – an approach that helps the understanding of the loadbearing behaviour of the stability system so that it is at least possible to check the results by means of simple manual calculations or single computer programs. The following calculation principles are extremely helpful for preliminary sizing or the design of stability systems. 2.2.3.1

General procedure for the calculations

When calculating the distribution of horizontal loads across the stability components, it is usually assumed that the floors distribute loads in the form of rigid plates. This assumption reduces the number of degrees of freedom per storey to three, i.e. two horizontal displacements and one rotation about a vertical axis. In the case of vertical stability components, the contribution of relatively soft components (e.g. columns) is neglected if the stiff components can provide the necessary stability alone. The distribution of the loads is determined according to the following scheme: 1. Reducing all stability components to one linear member with cross-sectional values changing for each storey. 2. Calculating the deformation of the rigid floor diaphragms as a result of the horizontal load. 3. Calculating the deformations of the individual stability components. 4. Calculating the internal forces for the individual stability components.

46


The method of analysis is considerably simplified if the stability components are arranged symmetrically on plan. The following cross-sectional values are assumed for the individual stability components: 1. Flexural stiffnesses EIy, EIx (kNm2) 2. Shear stiffnesses GAsy, GAsz (kN) 3. Torsional stiffness GIT (kNm2), made up of a) St. Venant torsional stiffness b) Bredt torsional stiffness 4. Warping stiffness ECM (kNm4) Asy and Asz designate the shear area. The following generally applies: ð 2 1 t dA w As Q

(8)

A

For rectangular cross-sections this simplifies to: 5 As w A (A w solid cross-section) 6

Table 2.7 Stability elements for skeleton structures Element

Load carried... in y-direction

in z-direction

via rotation about x-axis

A) Shear wall

Flexural stiffness EIz Shear stiffness GASy

– –

– –

B) Frame (girder)

Equivalent shear stiffness GA*Sy

–

–

C) Segmented shear wall

Equivalent flexural stiffness EI*z Equivalent shear stiffness GA*Sy

– –

– –

D) Open section


EIy GASz

(Torsional stiffness GIT) Warping stiffness ECM

E) Closed section


EIy GASz

Torsional stiffness GIT (Warping stiffness ECM)

F) Closed segmented section

Equivalent flexural stiffness EI*z EIy Equivalent shear stiffness GA*Sy GASz

Equivalent torsional stiffness GIT (Warping stiffness ECM)


47

The following can serve as stability elements (Table 2.7): – – – – – – –

Closed cross-sections Open cross-sections Plates Plates made up of precast concrete components Girders Frames Columns

The general arrangement on plan and the definition of the axes can be found in Table 2.7 and Fig. 2.30. The theoretical principles for systems made up of frames and plates are presented in [58– 60]. However, manual calculations are too complicated without applying simplifications. It is normal these days to use computer programs specially developed for calculating the service cores of high-rise buildings. Such programs perform the aforementioned calculation scheme. With only three degrees of freedom per storey, the computer processing time is minimal. The programs normally require the plates and cores with their cross-sectional values to be entered. The cross-sectional values can be determined by including an upstream program for calculating thin-wall cross-sections. Such a program included downstream calculates the bending and shear stresses in all the individual parts of the stability elements. Computer programs for designing the stability components are also available. In the case of thin-wall sections, it is appropriate to use programs for designing any reinforced concrete sections subjected to biaxial bending. However, such programs generally only allow the design for biaxial bending and not for torsion. Stability elements rectangular on plan (individual shear walls) can also be designed as columns.

Fig. 2.30

Plan of and general designations for building stability

48


Abandoning the idea of assuming that floor diaphragms are rigid increases the number of degrees of freedom considerably. The greater computing requirement is only justified in the case of especially soft floor diaphragms (e.g. large openings). In such cases it will be necessary to make use of universal linear-member programs for calculating three-dimensional frames, or FEM programs. And the disadvantage of grillage programs [61] is that stability elements positioned transverse to the loading direction cannot be taken into account. 2.2.3.2

Equations for rough preliminary design

The following simplifications are possible at the rough preliminary design stage: 1. The torsional stiffnesses of the stability components themselves can be ignored in the case of open sections provided the stiffness with respect to rotation is essentially provided by the warping stiffness of the total system. According to [50], where GIT 2 h J 0,25 (9) EIW the torsional stiffness GIT theoretically no longer has any influence (h w height of building). k2 w

2. Ignoring the warping stiffnesses ECM of several stability components is always possible if they are low compared to the total warping stiffness of the system. This is mostly the case with stability elements held apart by struts. The total warping stiffness EIw of a stability system is calculated as follows: EIw w

n X E CMi S Iyi y2i S Izi z2i

(10)

iw1

3. The shear deformation of beams is low compared to their bending deformation. Therefore, the shear stiffness of the stiffening plates can assumed to be infinitely large where several storeys are involved. Provided the stiffness and loading are constant over the height, the transverse distribution of the loads is then also constant over the height. However, comparative calculations indicate that the shear deformations of the walls in the case of stocky stability systems or in the lower storeys of tall buildings can lead to a considerable redistribution of shear forces. 4. The main axes of the stability elements are assumed to be parallel with or perpendicular to the loading direction. Torsional moments (Iyz) are not considered. The simplifications 1– 4 above mean that the equations according to [50] apply: Coordinates of the shear centre M0 of the total system: n P

y0 w

iw1

n P

Iyi yi

n P

iw1

, z0 w Iyi

Izi zi

iw1 n P

iw1

Izi


49

Load apportioning for bending action effects: qyj w

EIzj EIyj qy0 , qzj w n qz0 n P P EIzi EIyi

iw1

(11)

iw1

Load apportioning for torsion action effects: qyj w

Izj zj Iyj yj mx0 , qzj w mx0 Iw Iw

Warping resistance of total system: n X Iyi y2i S Izi z2i Iw w

(12)

(13)

iw1

Designations (see Fig. 2.30): qy0, qz0 mx0 Iyi Izi y j, z j n

horizontal load on total system torsion load about torsional axis of total system moment of inertia with respect to bending about y-axis of element i moment of inertia with respect to bending about z-axis of element i coordinates of shear centre of element j number of stability elements

Exploiting symmetries is appropriate for these equations. The calculation becomes especially easy when the load passes through the torsional axis. It is remarkable that with the stiffness distributed constantly over the height, the torsional axis is then only a vertical straight line when either only bending deformations or only shear deformations are entered. In the case of combined precast/in situ concrete systems, the torsional centres lie on a curved line. It therefore becomes clear that the equations given above can only apply to composite systems when the shear deformations are neglected. 2.2.3.3

Interaction of shear walls, shear walls with series of openings and frames

Where stability elements with different deformation behaviour, such as shear walls, segmented shear walls and frames, are designed to act together, then the relationship between their flexural and shear stiffnesses must be sensibly organised before the distribution of the horizontal loads can be calculated. Equivalent cross-sections for frames and segmented shear walls must be determined beforehand when the computer programs used for checking the stability of the building permit only solid plates or thin-wall cross-sections as stability elements. a) Segmented shear walls

The structurally equivalent shear wall is determined in such a way that its deformation under horizontal loads matches the deformation of the segmented shear wall as closely as possible. The calculation can be carried out in two steps as follows:

50


Fig. 2.31 Segmented shear wall and calculation models with continuous horizontal members (“rail” forces), and with FEM

1. Determination of the deflection at the top and further deflection by means of manual calculations or computer. The usual manual methods for segmented shear walls [50, 62, 63] are based on a method of calculation that replaces the single “rail” with a continuous row of lamellae (Fig. 2.31). Therefore, the flexural stiffness of a segmented shear wall lies between that of a solid shear wall (rigid anchors) and two separate shear walls. The magnitude of the rail’s moment of inertia is specified in [50] or [63] for a rail at floor level. A critical study of the use of an equivalent frame model for shear walls and high-rise building cores can be found in [81]. Special programs for plates now allow plane systems to be calculated on desktop computers. They permit almost all conceivable geometrical irregularities to be taken into account: – Openings – Holes – Individual linear-type elements – Different thicknesses Internal forces and design proposals can be output for all elements. 2. Determination of the unknown I* (moment of inertia) and A* (shear area) for the equivalent solid plate from equations (1) and (2) according to Fig. 2.32. b) Plates with large openings

Openings in shear walls are common in the lower storeys, i.e. in the zones with the highest shear forces. Such openings considerably diminish the stiffness of any shear wall. Fig. 2.33 shows an example of a shear wall with a large opening at ground floor level. The equivalent shear wall is divided into two areas with different cross-sectional values. The calculated cross-sectional values of the equivalent shear wall can be entered directly into a program for determining the horizontal load distribution.

51


Fig. 2.32

Segmented shear wall and associated equivalent shear wall

Fig. 2.33 Shear wall with large opening at ground floor level

c) Frames and girders

Frames and girders can be replaced by shear-equivalent plates when calculating the load distribution. The shear area of such plates is chosen so that the deflection at the top due to horizontal loads corresponds to that of the frame or girder (Fig. 2.34). The bending and strain deformation of the plate is set to zero, i.e. theoretically, the shear-equivalent plate has infinite flexural and strain stiffness.

Fig. 2.34 Equivalent shear wall for frame or girder

52


(More accurate calculations with FEM programs for plates and determination of equivalent cross-sectional areas according to Fig. 2.32). d) Three-dimensional systems

Manual calculations for three-dimensional stability elements with series of openings are only possible when a corresponding plane system can be found by exploiting symmetries. But everyday FEM programs should be used for general systems. One common case is the perforated hollow box. If this is symmetrical about its axes on plan, the bending action effects can be dealt with by means of a plane system. However, in the case of torsion action effects, a realistic model must be found that lies between the two extreme cases of – two U-sections, and – one closed hollow box-section (Fig. 2.35). As the closed hollow section yields far less under torsion loads than the two open sections, it is necessary to design the “rail” to be as stiff as possible in order to come as close as possible to the closed section. With a suitably stiff rail (rigid anchors), the warping stiffness can be neglected, although this represents an approximation. In this case the Bredt torsional resistance of the section can be calculated easily in two steps: 1. For a plate of thickness d with regular openings, an equivalent plate of thickness t* (I d) without openings is first determined based on the assumption that both plates exhibit the same shear stiffness (Fig. 2.36). In doing so, the system length l1 or h1 associated with the weaker cross-section can be reduced according to the principle of St. Venant if the depths of the “rail” and “post” cross-sections are very different. The equations given in Fig. 2.37 can be used for calculating the symmetrical case. An evaluation of the equations shows that the equivalent wall thickness is very low when either the rail depth or the width of the plates adjacent to the opening, i.e. the posts, is small.

Fig. 2.35 How the load is carried in a segmented hollow box


53

2. The Bredt torsional resistance is calculated as follows: 4A2 IT(Bredt) w P s

(14)

t

where A w (b - d) · (bl - d) s w length of one wall with constant thickness t Assuming that post shear forces are distributed continuously over the height, a general method of calculation was developed in [64] for perforated high-rise building service cores which also takes the warping stiffnesses into account. This method serves as a basis for a computer program [65] which also permits the coupling with other stability elements.

Fig. 2.36

Determining the equivalent thickness t* for a hollow-box wall with a series of openings

54

Fig. 2.37

2.2.3.4


Equivalent wall thickness of symmetrical hollow-box wall with a series of openings

Shear walls of precast concrete components

A shear wall made up of storey-high precast concrete components (Fig. 2.38) is not as stiff as a solid shear wall of the same size because displacements in the vertical joints can occur under horizontal loads. The prerequisite of course is that the horizontal joints possess sufficient shear resistance. We must distinguish between the following cases when determining the stiffness: a) The vertical joints are profiled and subsequently filled with grout. b) The vertical joints are smooth, the anchorage of the individual shear walls is achieved exclusively via the floors (see [36]). c) The shear walls are connected in the vertical joints by steel plates at individual points. The determination of the wall thickness for case c) is described in [82], and case a) is described in detail in [67].

Fig. 2.38

Shear wall assembled from precast concrete components


55

For reasons of economy, however, walls of precast concrete components are installed without grouting the joints afterwards – case b). During design, the anchorage effect of the floors is frequently totally neglected and therefore the wall reinforcement is overdesigned. The example below explains a method of calculation that takes the anchorage effect of the floors into account. There is no grout in the vertical joints of the wall shown in Fig. 2.39. The wall elements are supported on the floors and the horizontal joints are packed or grouted. However, the vertical loads are generally not sufficient to cancel out the tensile stresses in the horizontal joints, which means that it often becomes necessary to include longitudinal reinforcement at the edges of the plate to bridge over the horizontal joints. In this example the vertical load is ignored and the wall calculated for the given horizontal load only. The system acts like a multi-ply cantilever beam connected by individual anchors. A severely deformed zone develops in the vicinity of the anchors due to the transfer of the anchor shear forces into the two adjoining shear walls. According to the principle of St. Venant, this disruption decays at a distance from the point of transfer roughly equivalent to the depth of the floor diaphragm.

Fig. 2.39 Example of a shear wall assembled from precast concrete components

56


A computer program is used to design the shear wall. Fig. 2.39 shows the chosen element mesh. In the vicinity of the anchors, the element mesh is refined to square elements with a side length equal to the depth of the floor. The powerful elements used can also cope with the abrupt transition to the quite large elements shown here. The result indicates that only about one-third of the total moment at the base of the wall (1752 kNm) is transferred to the three plates in the form of bending moments. The axial forces transferred via the anchors into the outer plates form a couple that carries about two-thirds of the total moment. The maximum anchor force of 86.7 kN can be resisted by shear reinforcement in the floor slab. With knowledge of the horizontal deformations, an associated equivalent shear wall can now be determined easily according to the previous section (see Fig. 2.32): 0,198 20 124 w 2,33 m4 (37%) (3 1,084 s 4 0,519) 30000

I*

A*

0,198 20 122 w 0,32 m2 (26%) (0,519 s 0,354 1,084) 13000

The ratios of the cross-sectional values of the equivalent shear wall to the solid shear wall are given in brackets (h q b q d w 12.00 q 7.24 q 0.20 m; l w 6.33 m4; As w 1.21 m2). With a grouted vertical joint and an assumed joint stiffness of K w 4500 MN/m2, according to [67] the reduced moment of inertia and the shear area are as follows: Il w

6,33 0,93 4 w 4,45m ð70%Þ 2,88 7,24 2 1S 3 S 0,25 12,00

(15)

Alw 1,21m2 ð100%Þ If the anchor effect of the floor slab or the grouted vertical joint were to be totally ignored, the equivalent cross-sections would be the totals of the individual cross-sections: IL w

3 2,403 0,2 w 0,69 m4 ð11%Þ 12

AL w

3 2,40 0,2 w 1,20 m2 ð100%Þ 1,2


2.2.3.5

57

Example of horizontal load distribution

A five-storey building, stiffened by a central service core and two shear walls at the gable ends, will be used as an example (Fig. 2.40). Firstly, the shear walls will be assumed to be solid (a) and then by way of a comparison, openings at ground floor level will be introduced (b). In order to illustrate the differences in the horizontal load distribution, the lateral wind load will be reduced to a single point load W applied at the top of the building.

Fig. 2.40

Example of the distribution of a horizontal load

58


One further simplification is the assumption that the elements are arranged symmetrically about the axes, which means that the wind load does not cause any rotation. The calculation of the horizontal load distribution is carried out with a computer program. The storey height is 3 m, the thickness of the stiffening shear walls 0.25 m. Equivalent cross-sectional values were established beforehand for the wall with the opening according to Fig. 2.33. The increase in the shear force proportion in the outer shear walls in case a) as we proceed down the building is clear from the diagrams. If the shear deformations were to be ignored, then the shear force proportion would remain constant over the height. In case b), on the other hand, the weakening of the shear wall cross-sections at the bottom of the wall mean that almost the entire shear force is allocated to the core. 2.2.4

Verification of building stability

2.2.4.1

Stability analysis for stiffening cores and walls

When analysing the stability components of a precast concrete structure, according to section 2.2.2 the following loading cases must be investigated: – Wind – Earthquake – Out-of-plumb effects The combination of the loading cases is carried out according to DIN 1055-100. In doing so, the design situations “permanent and temporary” (wind and out-of-plumb) plus “earthquake” (earthquake, wind, out-of-plumb) must be examined. It may be necessary to consider horizontal forces resulting from residual foundation rotation (as a result of permanent eccentric core loading). The effect of the vertical loads on the deformed system gives rise to an additional horizontal action effect (Fig. 2.41). This is evaluated with a second-order theory calculation. The general procedure for the calculation accord-

Fig. 2.41

Calculation according to second-order theory


59

ing to second-order theory is based on determining the horizontal force distribution according to the previous section. 1. Distribution of the horizontal loads due to wind, earthquake, foundation rotation and eccentricity over the stability elements (shear walls, cores, frames). 2. Analysis of the individual stability elements under horizontal and vertical loads taking into account the horizontal deformation. Here, about 55–70 % of EII according to [50] or [68] can be assumed for EIII. The additional loads resulting from the second-order theory calculation are applied to each individual stability element; however, this ignores the influence of the coupling through the floor diaphragms for the additional loads. The building’s stability is verified by proving the stability of each individual stability component. The influence of foundation rotation is taken into account by elastic fixity in the ground. The torsion spring constant should be assumed to be as follows: cf w IF ce w IF

Ed pffiffiffiffiffiffi ([71], S. 527) 0,25 AF

(16)

where c@ torsion spring constant (MNm) IF moment of inertia of base area (m4) ce subgrade modulus (MN/m3) AF base area (m2) Ed coefficient of compressibility of soil for short-term loading (MN/m2). The analysis according to second-order theory can be omitted when it is established beforehand that the system is stable. DIN 1045-1 specifies stiffening criteria (previously lability coefficient or its inverse) to ease the assessment of the stability. These provide information on the yielding capacity of the stability components. However, strictly speaking, the use of these criteria is linked to further restrictive conditions as well as the assumption of rigid floor diaphragms: 1. The centroid axis on plan and the stiffness axis of the total stability system coincide (coaxiality). 2. Stability components have thin-wall cross-sections and their properties are constant over the height. 3. Vertical loads are identical in all storeys and run longitudinally along the centroid axis on plan. 4. All storeys are equal in height. 5. Foundation rotation is neglected. Even if not all of these conditions are met, a rough analysis of the building stability is still possible. In cases of doubt, however, a more accurate analysis will be needed.

60


Accordingly, the analysis according to second-order theory is unnecessary when the following applies: rffiffiffiffiffiffiffiffiffiffiffi 1 Ecm Ic j 1=ð0,2 S 0,1 mÞ for m J 3 hges FEd j 1=0,6 for m j 4 (17) Similarly, an assessment criterion for the rotational stability was derived in [69] and [50]. The following applies to systems with highly asymmetrical arrangements of stability elements or where torsional rotations cannot be neglected: vffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi vffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi u 1 u uPEcm Iv S 1 uPGcm IT t t hges FEd,j rj2 2,28 FEd,j rj2 j

j

j 1=ð0,2 S 0,1 mÞ for m J 3; j 1=0,6 for m j 4 With a uniform load distribution on plan, the value where ffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi r iw

(18) P j

FEd,j rj2 can be replaced by N i2,

i2x S i2y

The following applies for buildings rectangular on plan: .pffiffiffiffiffi i w d 12 O 0,289 d where m number of storeys height of loadbearing structure from top of foundation or a non-deformable rehtot ference plane rj distance of column j from shear centre of total system sum of design values of vertical loads, with gF w 1.0 FEd FEd,j design value of vertical load on column j, with gF w 1.0 Ecmlc sum of nominal flexural stiffnesses of all vertical stability components Ecmlv sum of nominal warping stiffnesses of all components providing stability against rotation which act in the direction considered GcmlT sum of torsional stiffnesses of all components providing stability against rotation (St. Venant torsional stiffness) Generalisations for torsion, such as considering mixed torsion and the calculations for non-coaxial systems, are described in [69, 70]. Foundation rotation can change the result of the stability criterion substantially. The stability condition has been extended in [68] depending on the moment of inertia of the base area and the subgrade modulus of the subsoil.


2.2.4.2

61

Stability analysis for columns and frames

In structures with adequate stability provided by cores and plates, the columns may be regarded as non-sway. DIN 1045-1 calls for the standard design of columns for the internal forces of the non-deformed system. A buckling analysis is required beyond a certain maximum slenderness. The model column method according to DIN 1045-1 is suitable for this analysis, provided the cross-section and axial force in a non-sway column remain constant in each storey. In more general cases it is often difficult to determine the buckling length required for the model column method. The reinforcement calculated as a result can be seriously overdesigned. Such cases are: – – – – – – –

Change of cross-section within free storey height Considerable load applications within storey height Vertical cantilevers, i.e. sway columns Staggered reinforcement Columns with elastic fixity at the base Suspended pinned-end columns Sway frames

In these cases the designer is recommended to carry out a buckling analysis on the deformed system (see also DAfStb booklet 525 [147]). As in this situation deformations, internal forces, the reinforcement required and the effective flexural stiffness are all mutually dependent and must be improved iteratively, calculation by computer program is usually the only option. Nowadays, programs are available for PCs which perform the buckling analysis according to second-order theory and carry out the design of the reinforced concrete sections, also for columns with biaxial bending, and include all transport, erection and final conditions. Unintentional eccentricity of the load application according to DIN 1045-1 section 8.6.4, with ea w aa1 l0 =2

(19)

must be considered in the buckling analysis. In doing so, it is important to investigate whether inclination of the total system compared to inclination of the individual column results in a less favourable value. When considering the total system, the deformed shape according to Fig. 2.42 must be considered. The experience to date shows that even with extremely slender and heavily loaded columns, creep usually has only a minor influence. A significant increase in the reinforcement required as a result of creep is only to be expected when a considerable portion of the bending moment critical for the design acts permanently (large permanent horizontal load; large intentional eccentricity of permanent vertical load). When carrying out a buckling analysis for sway systems according to second-order theory with the help of a computer program, the horizontal forces from pinned-end columns

62


Fig. 2.42 Different approaches to the deformation of stepped single-storey shed columns

or individual, very soft but it rigidly connected columns are taken into account automatically. Performing the deformation analysis for the total system avoids the shortcomings of the model column method for linear members and results in a much better assessment of the real situation (see also [72]). 2.2.5

2.2.5 Structural design of floor diaphragms

The individual elements of a suspended floor must be interconnected to form a floor diaphragm and connected to both the cores, which provide restraint, and the columns, which require restraint. In the case of composite plank floors and double-T floor units with structurally effective in situ concrete topping and shear reinforcement, this reinforcement with all the necessary connections can usually be laid in the concrete topping without any problems (Fig. 2.43). According to EC 2, even suspended floors made from prestressed hollow-core planks without shear reinforcement can achieve a plate effect by including a 5 cm deep concrete topping reinforced with welded mesh which is only connected to the loadbearing structure in the region of the perimeter and centre beams. However, where the floor diaphragm consists exclusively of precast concrete elements, these must be interconnected with a compression-resistant grout filling in the joints – apart from the fact that they must form a coherent planar element. The horizontal loads acting on the floor diaphragm are carried by truss action, with the ties necessary for this being provided by the longitudinal reinforcement in the joints or the perimeter members or by welding together the reinforcement forming the perimeter ties cast into the floor ele-

Fig. 2.43

Floor diaphragms with concrete topping


63

Fig. 2.44 Floor diaphragms assembled from precast concrete components, without concrete topping but with welded joints

ments during production. The disadvantage of the latter is that it can increase the number of different production positions substantially (Fig. 2.44). The longitudinal reinforcement in the joints acts as the tensile bending reinforcement for the plate or as the tension hangers of a truss model. The compressive forces in this truss are generally carried away diagonally via the joints. In order to transfer this shear in the plate it is sufficient when the joints can transfer shear forces in the longitudinal direction of the joint, acting like a hinge (Fig. 2.45a). This is achieved through the use of an appropriate shear key, in the case of very high loads by welding the edges of the joints together via cast-in steel items. If the joints also have to transfer load-distributing out-of-plane shear forces, then a shear key must be formed in both directions (Fig. 2.45b). Accommodating the ensuing horizontal expansion forces can likewise be achieved via the longitudinal reinforcement in the transverse joint (Fig. 2.46). Transferring thrust and shear forces in the diaphragm joints can also be via achieved looped reinforcement in the joints. However, this is a nuisance during production because then the side panels of the floor unit moulds are penetrated by the loops and, in addition, threading the longitudinal joint reinforcement through the loops on site is laborious. One remedy is to use specially developed looped wire rope connections which can transfer both in-plane and out-of-plane forces (see also section 3.3).

64

Fig. 2.45 Shear key joints (a) for in-plane shear forces (b) for in-plane and out-of-plane shear forces


Fig. 2.46 Structural effect of a grouted shear key

The horizontal component of the diagonal compressive forces resulting from the shear force action effect is carried via the floor diaphragm to the longitudinal reinforcement in the transverse joints. It is certainly the case that with floor diaphragms made up of individual precast concrete units, far more favourable loadbearing and deformation behaviour is achieved when instead of just one perimeter tie, longitudinal reinforcement is positioned in every joint, which of course must be suitably anchored in the perimeter member. Besides their function as tension hangers, i.e. as shear reinforcement (“links” or “stirrups”) in the floor diaphragm, the longitudinal reinforcement in the joints also has to accommodate the wind suction loads and the tie-back forces due to eccentricity. This is why they must be anchored in the outer columns or – with a structural grid offset internally – must incorporate these with loops. For in the end these have the task of holding the structure together adequately in the event of accidental loads (earthquake, explosion). The design of the floor diaphragm also essentially depends on whether it transfers the horizontal loads to the vertical stability elements (walls or cores) via compression or tension, and whether the force transfer is continuous over the entire depth of the diaphragm or merely concentrated over relatively narrow shear walls. The joints should be as narrow as possible to facilitate the transfer of the shear forces. However, they must be wide enough to accommodate the longitudinal reinforcement necessary, also at laps, and allow the (low-shrink) grout to be easily poured and compacted. Fig. 2.47 shows a precast concrete suspended floor construction with longitudinal joint reinforcement that is anchored in the columns. Other potential floor connections are described in [74]. There is no uniform concept for the design of the perimeter tie and the joint reinforcement. Truss action models are appropriate for their design [75]. In principle, when choos-

65


Fig. 2.47 Floor diaphragm without concrete topping but with screwed and lapped joints

ing a suitable truss action model, it must be ensured that the load can be carried with minimal deformations. The flow of the forces should therefore preferably take place via the relatively stiff struts. It is normal to carry all the tensile forces via a single tie around the perimeter (Fig. 2.48a). However, it is frequently impossible to incorporate this tie with the appropriate anchorages and corner details. It may well be better to distribute the tensile forces over several joints, as shown in Fig. 2.48b. The advantage of this solution is that the tying-back of the support reactions can be distributed over several places. In single-storey sheds it is frequently possible to omit the roof bracing completely because the horizontal forces are carried directly by the columns. If it is necessary to form a plate in the plane of the roof, this can either be achieved via the roof covering itself or via a truss, the chords of which are formed by two neighbouring rafters (see Figs 2.49 and 2.50). Where the roof covering is made from autoclaved aerated concrete panels, pumice concrete hollow-core planks [76] or trapezoidal profile metal sheeting, the approval documentation for these products contains the construction details necessary for achieving a plate effect with the individual elements.

66

Fig. 2.48


Truss action models for floor diaphragms with perimeter support

Fig. 2.49 Floor diaphragm formed by coupling beams together

Fig. 2.50 Zu¨blin House, roof plate over glass atrium formed by coupling two rafters together


2.2.6

67

Structural design of vertical stability elements

Walls and cores provided for stability purposes are mostly closely associated with the stairs, lifts and service shafts serving the various floors. In terms of construction, these vertical routes only require openings in the floors. Stair flights and landings can be supported, like the suspended floor elements, on the beams. However, as each floor constitutes the horizontal termination of a fire compartment and preventing the spread of fire from one storey to another is vital, the building regulations call for floor penetrations to be lined to achieve the appropriate fire resistance. Concrete walls are not the only answer here; it is possible to use lighter and less expensive materials (e.g. gypsum products or clay masonry). Nevertheless, concrete walls can also provide stability as well as fire protection. Service cores also frequently accommodate sanitary facilities and so concrete walls can also ensure the necessary sound insulation. So a range of different functions must be considered when planning a building core. During the planning it must be remembered that vertical shafts usually disrupt the modular coordination concept of a frame system. Apart from that, the construction and erection progress of a building is very severely affected by the method of building the walls required for the stability of the structure during construction as well as the final condition. It should always be remembered that every vertical shaft inevitably requires doors and openings. These reduce the stiffening effect of a shaft wall and, more importantly, every opening in a concrete wall disrupts its construction and hence the workflow. Therefore, the initial planning should consider only those concrete walls that extend undisturbed over the full height of the building. In the lift shaft that means generally three or four walls. With dog-leg stairs, which are typically required with storey heights of 3– 4 m, that almost always means the two longitudinal sides of the stair shaft and often also one end wall adjacent to the intermediate landings. Compared to four-sided, closed boxes, open U-shaped shafts have a comparatively low torsional stiffness (see section 2.2.3.3). Where such a core is located eccentrically on plan, designers are therefore recommended to also include the fourth side, which contains the openings, in the stability considerations, provided the torsional stiffness needs to be taken into account for the stability of the building. However, in such a case, the “rails” above and below and the “posts” to either side of each opening must be of sufficient size. This is where the designer has to weigh up the constructional difficulties against the gain in stiffness (Fig. 2.52). Fig. 2.51 shows a closed in situ concrete core for an industrial building made from precast concrete components during construction. Whenever possible, intermediate walls in hollow boxes, which hardly increase the bending or torsional stiffness, should not be constructed in concrete (Fig. 2.53). Walls and cores required for stability can be built of in situ concrete or from precast concrete elements (Fig. 2.54). Cores with in situ concrete walls are mostly built using climbing formwork (slipforming for very tall buildings only) (Fig. 2.55). With such forms of construction, brackets and corbels projecting from the wall surface should be avoided if at all possible. Any such items required should be attached afterwards. The tension reinforcement is installed by means of screw couplers, and reinforcement in the form of dou-

68


Fig. 2.51 In situ concrete core built before the erection of the frame

Fig. 2.52

Service shafts as stiffening cores

Fig. 2.54

Shear walls

Fig. 2.53 walls

Stiffening core with masonry internal

69


Fig. 2.55

In situ concrete core, constructed with climbing formwork

Fig. 2.56

Beam corbel fitted into pockets

Fig. 2.57

Fig. 2.58

Corbels cast on afterwards

Connecting a landing slab to a stair shaft wall

ble-headed studs is ideal because of the rotating head (Fig. 2.57). The best solution is to support beams or floor slabs in pockets (Fig. 2.56). Internal landing slabs can be attached to stair shaft walls via steel sections and grouted joints (Fig. 2.58). Shear walls made from precast concrete elements according to Fig. 2.54c are particularly common in large-panel construction. Storey-high precast concrete elements should be

70


Fig. 2.59 Different horizontal reinforcement arrangements for shear walls [73]

used in such designs. The shear forces in the joints must be analysed. The reinforcement designed to accommodate the splitting of the shear force into a horizontal tension component and a diagonal compression component may be concentrated at the level of the floor slabs according to Fig. 2.59 for walls whose total width is greater than the storey height. Shear reinforcement (e.g. loops) distributed over the height of the joint is recommended in [73] for vertical wall joints between precast concrete elements at the corners (L-, T-, U-cross-sections on plan) (see Fig. 2.60). Reinforcement concentrated at the level of the floor diaphragms only cannot prevent the joint opening. The horizontal joints in walls of precast concrete components are primarily loaded in compression. Transferring the shear forces is then generally guaranteed via friction. Shear keys such as those shown in Fig. 2.61 will be necessary in certain cases. Any tensile forces that occur (see section 2.2.3.4) can be accommodated by welding (Fig. 2.62), screw couplers (Fig. 2.63) or approved built-in items plus screw couplers (Fig. 2.64). Stiffening according to Figs 2.65 and 2.66 is employed for frame structures that are conceived as fully prefabricated systems. Further design and construction advice for shear walls made from precast concrete components can be found in [77] and other publications. Using frame systems (Fig. 2.54f) to provide stability is generally only worth considering for special cases. One example is the structural carcass for a paper mill (Fig. 2.68). The rigid connection between beam and column (Detail A) is achieved with the help of screw couplers that are subsequently pressed tight. Screw couplers with metric or conical threads are now available. The shear force is transferred via a corbel, or rather via the grout filling in the gap between beam and column. Detail B shows the hinged but laterally restrained support for the roof beam at the column. Remarkable here is the separate transfer of the vertical support reaction via the cast-in neoprene bearing pad and the horizontal


71

Fig. 2.60 Deformation of joints between longitudi- Fig. 2.61 Core wall with horizontal shear key nal and transverse stability girders with different joints for high shear and low axial forces reinforcement arrangements [73]

support reaction via the cast-in shear connectors. The neoprene pad allows the vertical support reaction to be distributed over a large area. Similar designs are described in [78–80]. Whether stiffening walls or cores are to be made from precast concrete elements or in situ concrete should be clarified with the contractor at an early stage because this has a decisive influence on the sequence of planning and building operations. Walls of precast concrete elements are generally somewhat more complicated and require more input during planning and therefore require an appropriate lead time for the technical work, which, however, can be recouped by the faster erection. On the other hand, in situ concrete cores can, or rather must, be built prior to the arrival of the precast concrete elements on site, i.e. at the same time as the elements are being produced in the factory. Whether this is possible is not only a question of the separation into different trades, but also a question of the time of year in which the building work can take place. Such decisions can only be reached on an individual basis taking into account costs and timetables.

72

Fig. 2.62

2.2.7


Welded joint between wall elements

Fig. 2.63 ments

Screwed joint between wall ele-

Design of perimeter ties to DIN 1045-1

According to DIN 1045-1 section 13.12, various ties must be provided according to Fig. 2.67 in order to a) limit local damage as a result of accidental actions such as impact or explosion; b) enable alternative load paths in the event of local damage. To achieve this in precast concrete construction, internal ties plus horizontal column and wall ties are required as well as perimeter ties. The characteristic strength of the steel fyk may be utilised to the full in the design of tie cross-sections. In addition, the designer may allocate existing reinforcement provided for normal action effects (according to section 2.2.2) to the perimeter tie.


Fig. 2.64 Joint between wall elements with (a) dowels plus grout, and (b) special cast-in wall starter units PSK (Peikkor)

73

74


Fig. 2.65 Stair shaft assembled from precast concrete components (production building for Siemens AG; contractor: DYWIDAG)

Fig. 2.66 Core assembled from precast concrete walls (Backnang Grammar School; contractor: Zu¨blin)


75

Fig. 2.67 Ties for accidental actions to DIN 1045-1

First of all, every floor level must include a continuous perimeter tie no more than 1.2 m from the edge. This should be able to accommodate a tensile force of FEd w li 10 ½kN I 70 ½kN where li is the span in [m] of the end bay of the floor diaphragm perpendicular to the perimeter tie being considered. Joints between reinforcing bars can be lapped or welded. Laps should be designed with a length of ls w 2 · lb and should be secured with shear reinforcement (links, U-bars, etc.). Internal ties must be provided in two directions at 90h to each other and at their ends they must be structurally connected to the perimeter tie. They should be able to accommodate a tensile force of FEd w 20 ½kN=m In floors in which the perimeter ties are positioned in the joints between the precast concrete elements, a minimum force per joint amounting to FEd w

ðl1 S l2 Þ 20 ½kN I 70 ½kN 2

should be assumed (l1, l2 in [m], see Fig. 2.67). Perimeter columns and external walls must be connected at every floor level; design tensile force FEd w 10 kN/m of façade, although the maximum force per column does not have to exceed FEd w 150 kN. Corner columns should be anchored in two directions, and here the outer perimeter tie may form part of the anchorage. In the case of buildings built using large panels and having five or more storeys, the walls must also be interconnected with vertical ties in order to prevent a floor collapsing in the case of the failure of the wall below, e.g. due to a local explosion. The perimeter tie should form part of a system that “bridges over” the damaged area. These ties should extend the full height of the building and in the damaged condition be able to accommodate the design value of the load acting on the floor immediately above the failed wall.

76


(a) Erection

(b) Section

(c) Structural System

Fig. 2.68

Stability provided by frames (Holtzmann paper mill; contractor: Zu¨blin)


(d) Detail A: frame corner

(d) Detail B: beam support

Fig. 2.68

contd.

77

78

2.3


Loadbearing elements

The design of structural precast concrete elements is essentially determined by the production methods. The principal dimensions are limited by the transport restrictions described in section 2.1.3. 2.3.1

Suspended floor elements

There are many precast concrete floor systems. Those described on the following pages have proved to be the most economic and/or most flexible. Fig. 2.69 shows the most common precast concrete floor systems, and the cross-sectional dimensions of solid and hollow floor units are given in Fig. 2.72 (p. 80). 2.3.1.1

Hollow-core slabs

The hollow-core slab is one of the most economic precast concrete floor systems, provided it can be produced in sufficient quantities in order to take advantage of its fully automated production. The circular, oval or even rectangular voids save materials and weight – up to 40 % compared with solid concrete slabs. We distinguish between prestressed and conventionally reinforced hollow-core slabs. In the prestressed hollow-core slab (Fig. 2.70) the reinforcement is exclusively in the form of pretensioned strands. The slabs are manufactured in prestressing beds more than 100 m long using slip-formers or extruders, which at the same time form the mould and perform the tasks of distributing and compacting the concrete. Concrete strengths of up to 60 N/mm2 can be achieved with these methods. After curing, the individual floor units are cut from the long ribbon with mechanical saws (see also section 4.1). This form of production only allows the reinforcement to be prestressed in the longitudinal direction, which means that prestressed hollow-core slabs require a National Technical Approval awarded by the DIBt if they are to be used in Germany. As can be seen from the table

Fig. 2.69

Precast concrete floors

2.3 Loadbearing elements

79

Fig. 2.70 Prestressed hollow-core slabs; top: range of products; bottom: erection of hollow-core slabs (Fachvereinigung Spannbeton-Fertigdecken e.V.)

in Fig. 2.70, hollow-core slabs are available in various depths. Spans of up to 18 m can be achieved with slabs approx. 40 cm deep. The standard width is 1.20 m. The highly eccentric prestressing causes upward creep in the slabs and the different deformations of the individual elements can lead to considerable problems at joints, particularly with the smaller depths. This aspect should be tracked during storage of the elements. The lack of space in the joints between the individual elements can cause problems when attempting to use the slab as a horizontal diaphragm for stability purposes, which is often necessary. This aspect requires careful planning. Certain voids can be partly filled at the supports in order to accommodate large in-plane loads, with the reinforcing bars then functioning as “dowels” between adjacent bays (Fig. 2.71).

80


Fig. 2.71 Special solution for forming a floor diaphragm with prestressed hollowcore slabs

Fig. 2.72

Cross-sections of floor slabs (Fachvereinigung Deutscher Betonfertigteilbau e.V.)

The omission of conventional reinforcement because of the method of production means that the tensile strength of the concrete has to be included in the calculation to some extent in order to achieve the necessary shear strength, especially at the supports, and to distribute the loads in the transverse direction. This is particularly the case when supporting the elements on a “soft” support (e.g. steel beam). Beam deflections give rise to transverse tensile stresses. The first studies on this have already been carried out [95] (see also [95-1]).

Conventionally reinforced hollow-core slabs are produced in widths up to 2.50 m on steel pallets in the desired lengths, generally in a special plant in which augers push the concrete through a rectangular die matching the dimensions and geometry of the slab cross-section (see also section 4.1). Both longitudinal and transverse reinforcement, including shear links, is possible. The slabs can essentially be designed according to DIN 1045-1 and do not necessarily need a National Technical Approval. Principles for the


81

structural design of conventionally reinforced precast concrete floor slabs can be found in [94], where these differ from DIN 1045-1. Slab depths are generally between 14 and 20 cm in order to achieve spans of 6–7 m for loads of 5 kN/m2. Spans of up to 10 m are possible with a slab depth of 30 cm. The longitudinal edges are cast with shear keys where necessary so that in-plane and out-of-plane forces can be transferred across the joints between elements. Hollow-core slabs can always be erected without the need for any temporary propping. 2.3.1.2

Ribbed slabs

Ribbed slabs, either conventionally reinforced or prestressed, will be required for higher loads and longer spans. The double-T slab is well established in the marketplace. Conventionally reinforced versions are produced in long moulds, prestressed versions on prestressing beds. The units are manufactured in widths up to 3 m, depths of 70 –80 cm and lengths up to 16 m (Fig. 2.73). The webs (i.e. the ribs), usually 1.20 m apart, have sides that slope at an angle of 1:20 so that the elements can be easily lifted out of the rigid moulds after the concrete has hardened. The side panels of the moulds can be adjusted to suit the respective slab width. The units are generally produced with a 6 cm deep flange which serves as permanent formwork for the subsequent in situ concrete topping added on site. The reinforcement required for creating the diaphragm effect is laid in this in situ topping.

Single-T slabs (i.e. T-beams) are generally only found in the form of trimmers in floor systems with double-T elements. However, they have already been used successfully as the vertical wall elements for a high-bay racking warehouse (Fig. 2.74). A variation on the double-T element is the inverted channel section unit (Fig. 2.75), which is used for heavier point loads or where the floor element width matches the column grid. In the form shown in Fig. 2.69f this type of floor slab has the disadvantage that a dropside or otherwise movable panel is required for the outer parts of the rib moulds. The larger span of the unit in the transverse direction calls for a flange at least 12 cm deep and more reinforcement than the double-T elements. Where the ribs are fully notched at the supports, a rib on the edge of an element is also less favourable than the rib of a double-T element connected to the flange on both sides. 2.3.1.3

Composite plank floors

Composite plank floors have been used extensively since the early 1980s [84] and today they represent the most popular flooring system in Germany. These are solid concrete floors once they are completed, with a 5–7 cm deep precast concrete plank containing the bottom reinforcement necessary for structural purposes and acting as permanent formwork for the in situ concrete topping. In order to be able to handle these thin elements, lattice beams made up of rigid reinforcement protrude from the top of the precast plank. The top chord of each lattice beam serves as a compression member in the tempor-

82


Fig. 2.73 Cross-sectional values of double-T floor units (Fachvereinigung Deutscher Betonfertigteilbau e.V.)

ary erection condition, the two bottom chords can be included in the tension reinforcement required for structural purposes. DIBt National Technical Approvals are available for various types of lattice beam. The floors are designed according to DIN 1045-1. The diagonals in the lattice beams and the rough top side of the plank guarantee an adequate bond with the in situ concrete topping, which means that the floor can be designed like a solid slab cast in one operation. A continuity effect in the slab can be achieved in a simple way by adding top reinforcement to the lattice beams on site. It is also easy to lay any additional reinforcement that might be needed to achieve a horizontal diaphragm effect.

83


Fig. 2.74 High-bay racking warehouse made from precast concrete T-beams (Zu¨blin system)

Fig. 2.75 Multi-storey car park with inverted channel section floor units (contractor: Zu¨blin)

A discussion surrounding the FEM design of floor slabs spanning in two directions can be found in [101] and [103]. According to these reports, FEM can also be used for precast concrete floor systems provided there is no joint within the torsion zone (0.3 · Lmin) (or the reinforcement for the transverse direction is laid in the in situ concrete topping) and the depth of the joint does not exceed one-third of the total slab depth. The reinforcement from a design for an in situ concrete slab can be used here provided the span reinforcement transverse to the span of the precast concrete elements is continuous across the floor

84


slab and is calculated according to the ratio of the different structural depths. The bar diameter should not exceed 14 mm and the reinforcement required for flexural tension should not be greater than 10 cm2/m. The introduction of the new DIN 1045-1 means that the permissible shear stress is currently still limited to 0.25 · VRd,max. It is now also possible to construct composite plank floors supported on individual columns (i.e. flat slab with no floor beams). Special lattice beams are required to resist punching shear. Where composite plank floors, like hollow-core slabs, are required to span up to 5 m without any temporary propping during erection, then the single-bar top chord of the lattice beam can be replaced by a channel-shaped, buckling-resistant sheet steel section that is filled with concrete at the same time as casting the precast concrete plank (Fig. 2.77). This type of suspended floor is especially economic with high storey heights if the extra cost of the lattice beam is lower than the cost of providing temporary propping. This system (trade name “Montaquick”) is also approved for non-static loads [96]. Erection and concreting of the floors without temporary propping can be achieved by using prestressed floor elements. To do this, the concrete plank is pretensioned with wires positioned almost in the centre of the depth. Spans of approx. 8 m during erection are then possible (Fig. 2.78) with prestressed concrete planks about 8–10 cm deep. Thermal expansion in the longitudinal direction would lead to restraint forces when using lattice beams. Therefore, shear reinforcement can only be achieved with the help of additional shear links. The shear connection between precast and in situ concrete is achieved by way of the bond between the two, although a positive form of shear connection should be provided at the supports. The use of prestressed composite plank floors is regulated exclusively via DIBt National Technical Approvals. A T-beam slab for longer spans is achieved by using conventionally reinforced or prestressed precast concrete beams, with shear connectors protruding from the top, to support composite plank floors spanning at 90h to these (Fig. 2.79). This creates a horizontal diaphragm as well as a continuity effect in the transverse direction.

Fig. 2.76 Two-way-spanning composite plank floor with punching shear reinforcement at columns (flat slab)

Fig. 2.77 Lattice beam for stable composite plank floor during erection (Montaquick, KaiserOmnia systems)

85


Fig. 2.78 Prestressed floor units (Scha¨tz Spandec system)

2.3.2

Floor and roof beams

2.3.2.1

Floor beams

A rectangular cross-section is the simplest form of downstand beam (Fig. 2.79a). However, this cross-sectional form is not the best for precasting because it requires dropside moulds. So this beam shape should only be used when it is required for architectural reasons or to solve connection problems in the interior fitting-out. Generally, and for purlins (Fig. 2.79b) almost exclusively, we therefore use trapezoidal cross-sections where the vertical sides slope outwards at an angle of 1:10 or 1:20, which are easier to lift out of a rigid mould. The bottom arrises are cast with a 10 mm chamfer. To reduce the depth of the floor construction, rectangular downstand beams are frequently cast with a continuous boot on both sides at the bottom (Fig. 2.79c) to support the floor units. This is not the optimum beam form for precasting but in many cases no other options are suitable. It requires a relatively elaborate mould mechanism that is only worthwhile for large batches or standardised products. In addition, the incoming floor loads must be “suspended” within the beam. And achieving a flawless surface finish on the boot often requires additional work. The boot should not be smaller than 20 q 20 cm in cross-section in order to guarantee an adequate bearing for the floor elements and sufficient anchorage for the reinforcement in the boot and the floor units taking into account the inevitable tolerances (see section 2.6.2). In the frame system according to [88] and [97] (Fig. 2.80, p. 90), an inverted channel section unit is used as a downstand beam. This somewhat more costly beam form is only worthwhile when considered in conjunction with the total system. The webs of the double-T floor units used here are notched over their full depth so that they can be laid on top of the inverted channel beams. The loads are therefore applied to the top of the latter and the overall depth of the floor construction is no greater than that of a beam with continuous boot. The suspension reinforcement is now located in the webs of the double-T floor units and forms part of the longitudinal reinforcement in the webs (see section 2.6.2). Corbels on both sides of the columns support the beam webs, which are long enough to reach from one column grid-line to the next. Even when used as perimeter beams,

86


(a) Fig. 2.79 Cross-sections for beams, downstand beams and wall panels (Fachvereinigung Deutscher Betonfertigteilbau e.V.)


(b) Fig. 2.79

contd.

87

88


(c) Fig. 2.79

contd.


89

(d) Fig. 2.79

contd.

the torsion in the webs is minimal, and depending on the position of the grid with respect to the façade, complete or half perimeter beams can be used. Building services can be routed in the voids between the webs.

90


Fig. 2.80 Frame system (6M system, Zu¨blin)

2.3.2.2

Roof beams

The most economic cross-sectional form for roof beams is the T-section with parallel flanges (Figs 2.81a and b). In the past there were attempts to minimise the material in such beams by using duopitch forms and keeping the web thickness down to the permissible minimum. This led to I-beams with additional bottom flanges and in some cases web haunches (Fig. 2.82a). These days, we prefer to cast such roof beams in long moulds, often prestressing beds, in some cases in long lines, one behind the other. Only in the case of long-span roof beams (i 25 m) is a wider bottom flange to accommodate the reinforcement unavoidable. Creating a slope for the roof covering is achieved either by placing the roof beams at the appropriate pitch or by forming notches of different depths in the purlins (Figs 2.82c and d). If a duopitch roof beam is unavoidable, then at least the underside of the top flange should be kept level wherever possible (Fig. 2.82b). Web haunches should generally be avoided because they require practically a doubling of the mould side panels over most of the length of the roof beam. Openings in the webs of downstand floor beams and roof beams are generally required for routing the building services. Such openings must be taken into account in the planning right from the start. Devising a certain basic system for each building is recommended so that every beam of a certain type can be based on the system (Fig. 2.83). The reader is referred to [190] for the design of beam webs with openings for services. Fig. 2.84 shows a two-bay single-storey shed. The spacing of the rafters is chosen to suit the roof covering, which is normally of trapezoidal profile metal sheeting or autoclaved aerated concrete panels. This results in economic rafter spacings of 5.0 –7.5 m. The economic span for such rafters is 12–24 m, but rafters spanning 40 m have been produced on occasions (Fig. 2.85, p. 94).


(a) Fig. 2.81

Roof beam cross-sections (Fachvereinigung Deutscher Betonfertigteilbau e.V.)

91

92


(b) Fig. 2.81

contd.

93


Fig. 2.82

Different roof pitch solutions

Fig. 2.84

Two-bay single-storey sheds

Fig. 2.83

Beam web openings, standardised

94

Fig. 2.85

2.3.3


Transporting a roof beam with a span of 40 m (contractor: Bremer)

Columns

The standard cross-section for precast concrete columns for single-storey sheds is rectangular, whereas the square form is generally preferred for multi-storey buildings (Fig. 2.86), where a constant cross-section throughout all storeys is the best solution in order to achieve uniform support and connection details, especially with respect to the interior fitting-out. Multi-storey buildings up to five storeys high are built with continuous columns. However, such long columns should not be too slender because their flexibility can then cause serious problems during transport and erection. The standard cross-section for columns in conventional buildings is 40 q 40 cm. In taller buildings with spliced columns, the splices should be offset in different storeys in order to improve stability during erection. Columns with a circular cross-section are also possible. However, if these are cast in vertical moulds, then only storey-high columns are possible, which adds up to a considerable number of splices. Circular columns can also be cast horizontally as hollow columns with very high concrete strengths using the spun concrete method, although such a form of production requires special facilities [93]. The best solution in terms of mould design is when corbels are positioned on two opposite sides (Fig. 2.87). A third corbel on the top during production is also possible. Corbels on four sides are to be recommended in exceptional cases only because of their difficult production. One example of this is the University of Riyadh [91, 92], which was designed so rigorously as a precast concrete structure that all the columns could be produced in the same type of mould. A double steel mould with an automatic dropside mechanism was developed according to a uniform scheme for the 2600 columns (Fig. 2.88). This enabled corbels to be positioned on all four sides where necessary. In the early years of school and university building, when unidirectional structural systems were still the goal, peripheral or annular corbels were designed so that there would be support for beams in both directions (see [3]). This type of corbel is very awkward in terms of mould and reinforcement and should be avoided. More recently, there have been more and more attempts to create concrete façades with special architectural effects again, which are economic when the

95


Fig. 2.86

Column cross-sections (Fachvereinigung Deutscher Betonfertigteilbau e.V.)

Fig. 2.87

Positioning of corbels

concrete also has a loadbearing function. This leads to architectural façade columns, e.g. as in the case of Zu¨blin House. A relatively elaborate mould was required for this project. But thanks to the simple and clear concept of the building, it was possible to produce all the façade columns in one type of mould, which included the “lugs” for three upper floors (Fig. 2.89) so that the columns only had to be spliced once or twice. Each mould was used

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Fig. 2.88 Column types for the University of Riyadh system [92]: (a) columns for one, two and three storeys (b) reinforcement (c) corbel positions

more than 100 times. These two examples, the University of Riyadh and Zu¨blin House, show how a project-related breakdown into prefabricated parts can certainly be sensible and economic, even though these differ from the planning principles for industrialised building systems in general. More and more storey-high precast concrete columns are being used, especially for composite precast/in situ concrete structures. One reason for this is the faster construction time for high-rise buildings, for instance, with the column splices in the form of simple butt joints according to section 3.1.1. A steel plate is usually specified to cope with the

Fig. 2.89

Standard column for Zu¨blin House; spring-loaded, “breathing“ column mould [94]


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Fig. 2.90 Butt joint between precast concrete columns for Triangel Tower in Cologne (contractor: Zu¨blin)

high loads normally encountered in high-rise buildings. The force transfer at the level of the suspended floors requires particular attention (Fig. 2.90) [104 –106]. 2.3.4

Walls

The precast concrete wall is the characteristic loadbearing element of large-panel construction (see [90, 98]). This section looks at internal walls only; external walls are dealt with in section 2.4 “Precast concrete façades”. According to DIN 1045-1, a minimum thickness of 8 cm is adequate for loadbearing precast concrete walls in conjunction with continuous floor slabs. However, the wall thickness is generally governed by the minimum bearing dimension required for the floor elements. Internal walls are therefore between 14 and 20 cm thick (see Fig. 2.79d). Furthermore, it is sound insulation and structural fire protection requirements that are the main criteria for internal walls. A 14 cm thick concrete wall ensures adequate sound insulation. This same wall thickness is also adequate for a fire wall or a F 90 fire resistance rating. In addition, concrete internal walls also help to achieve summertime thermal performance requirements thanks to their good thermal mass.

Composite precast/in situ concrete walls (Fig. 2.91) represent an ideal combination of the advantages of both types of construction. The expensive formwork operations are transferred to the factory and the finished, cast wall is monolithic with a smooth surface both sides. Such walls have in the meantime secured a significant market share for themselves. They are used in almost all buildings and have also been used for civil engineering works [107]. Owing to their fast erection, such composite walls are especially suitable for those walls that would require formwork on one side only when cast on site (e.g. building against existing works, etc.). However, the design loads should not be too large because there is a limit to the amount of reinforcement that can be placed between the precast “leaves”. The in situ concrete should be at least 10 cm thick, which together with the

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Fig. 2.92 Use of composite precast/in situ concrete wall for tall walls (SysproPART system)

Fig. 2.91 Composite precast/in situ concrete wall with lattice beams and in situ concrete fill (SysproPART system)

two outer precast leaves each 6 cm thick, results in a minimum wall thickness of 22 cm. Slimmer walls are possible in certain circumstances but the concreting operations must then be planned in great detail. Composite walls are also useful where, for tall walls, the weights of precast concrete panels would exceed the lifting capacities of the cranes available (Fig. 2.92). One important application for composite walls is in basements. Initially used only for internal walls, they are being increasingly used for external walls, and also for impermeable

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concrete basements. The DAfStb directive covering impermeable concrete [108, 109] specifically mentions this system. A key advantage is that any cracking is restricted to the joints between the precast concrete wall elements. But in the end, good-quality workmanship at the joints and a minimum in situ concrete thickness of 20 cm are critical to their impermeability. Currently there are still misgivings in some circles concerning the discrepancy between theory and practice because the risk of flaws and the quality demands are very high [110]. 2.3.5

Foundations

Foundations are heavy and therefore usually cast on site. Nevertheless, precast concrete foundations are also feasible. Fig. 2.93 illustrates the “evolution” of this form of construction. The pad foundation with separate smooth-sided pocket on top, which was common for many years, has been replaced by the true pocket foundation (Fig. 2.95) with a pocket formed in the foundation itself, which is more economic [99]. Foundations can therefore be shallower and the separate pocket, whose expensive forming and reinforcing processes always had to be carried out in a separate operation, is no longer necessary. However, a column inserted into a pocket always requires an adequate key between the base of the column and the walls of the pocket so that the axial forces can be transferred to the foundation via skin friction. It is relatively easy to fix trapezoidal battens to a drop-

Fig. 2.93

Foundation types

Fig. 2.94

Profiled column base

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Fig. 2.95 Column pocket formwork: a) corrugated sheet metal tube, b) formwork box with special corner fittings

side column mould (Fig. 2.94). And pockets are in many cases formed with permanent formwork in the form of a corrugated square sheet metal tube (Fig. 2.95a). Mould boxes with special unplasticised PVC corner connections and clamping bolts are available. The bolts are undone for demoulding so that the four mould sides can be separated from the concrete with a light blow from a hammer (Fig. 2.95b). Which mould form is more economical depends on the particular costs, the particular situation. Precast concrete columns complete with precast concrete foundation already attached have been available for some time (Figs 2.93 and 2.96). This overcomes the need for a column–foundation connection detail and the foundation can be produced in the works together with the column. Once on site, the precast concrete element is lowered onto a layer of blinding and aligned with steel shims before grout is pumped underneath to create a structural bond between the underside of the foundation and the subsoil. Vertical pipes must be provided in the foundation to ensure good distribution of the grout and prevent air pockets. The system leads to a further shortening of the construction time and to foundations at a shallower depth. The disadvantage of the system is the bulkiness of the precast concrete element (column S foundation) and the ensuing economic transport. Transport restrictions mean that in one direction the foundation can be no more than 3 m wide. However, in situ concrete can be used to increase the size of a foundation on site. In the University of Riyadh project, the connections – as is very common in the USA – were designed by American engineers according to the principles common to structural steelwork. The columns were fitted with cast-in steel base plates that were then screwed to holding-down bolts cast into the foundations (Figs 2.93 and 2.97). This system is becoming more and more popular in Germany, too. The disadvantage of a high steel content within the connection is balanced by the advantages of the easy fabrication of the base plate, the easy casting of the foundation (no pocket required, etc.) and the relatively shallow structural depth of the foundation. This

2.4 Precast concrete façades

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Fig. 2.96 Precast concrete column complete with precast concrete foundation (contractor: Bachl) Fig. 2.97 Column base detail for system shown in Fig. 2.88 [91]

can be very advantageous on sites where, for example, there is a high ground water table. With moderate column loads, the steel plate can be omitted completely and separate anchorage elements used instead. It is vital to cast the holding-down bolts into the foundation as accurately as possible using a template and to protect these against damage until column erection begins. Tolerances of approx. e5 mm are possible. The system allows the full column fixity required for the structural design to be achieved if required, but also merely temporary erection fixity. These days, it is no longer common to pack the joint; grout is used instead, which is pumped in via tubes, which also prevent air pockets. 2.4

Precast concrete façades

In contrast to the design of structural elements, where the manufacturing requirements are the primary concern, the design of components for the external envelope of a building is mainly determined by the demands of architecture and building physics. As the building envelope is in this case not a homogeneous surface, but rather an assembly of individual

102


elements, great attention must be paid to the appearance, functions and constructional aspects of the joints and fixings. Façades made up of precast concrete elements are dealt with in general in [7, 111, 112, 134]; more recent architectural developments are presented in [9], for example. After many years of steel-and-glass architecture, we are now witnessing a revival of “architectural” façades. The precast concrete elements used – in contrast to conventional concrete façades – are required exclusively for appearance purposes and reveal the diverse design options available with precast concrete. Another interesting development is the use of building envelopes made from small-format, thin façade panels of glass fibre-reinforced high-strength concrete. This section will first deal with conventional façades made from precast concrete elements, the requirements they must satisfy and the details, before taking a look at new architectural developments. 2.4.1

Environmental influences and the requirements of building physics

The external climatic influences that affect façades are, first and foremost, solar radiation and rain together with wind pressure and outside temperatures. Facing these on the inside are room temperatures, the humidity of the interior air and water vapour pressure (Fig. 2.98). So in order to repel or attenuate these various influences, a façade must therefore function as reflective layer, rain screen, airtight membrane, thermal insulation, thermal mass, surface condensation absorber and vapour barrier all at the same time [113, 135]. With the exception of thermal insulation, concrete is an ideal material for satisfying all these requirements. Furthermore, concrete façades are good for sound insulation and fire protection, too, and their high strength can be exploited for loadbearing purposes. Factory production in particular offers further options that enable concrete to be constructed in virtually any shape, with a huge variety of surface finishes and colours, possi-

Fig. 2.98

Climate factors and wall functions [113]


103

bly also with facing leaves of brickwork, stone or metal. So with all these advantages it is not surprising that precast concrete façades are used not only on precast concrete buildings, but also to clad in situ concrete and structural steelwork as well. Concrete façades are normally found in the form of three-ply sandwich panels, with facing leaf, thermal insulation core and loadbearing leaf, which are manufactured in one operation and erected as complete units (Fig. 2.99a). The layer of thermal insulation, usually made from PS or PU rigid foam boards, should preferably be positioned closer to the outer side of the wall panel. Conventional precast concrete sandwich panels with the layer of insulation concealed behind render (Fig. 2.99b), or a central layer of insulation and thin concrete facing leaf (Fig. 2.99a), or a rendered, single-leaf façade of dense lightweight concrete (Fig. 2.99d) represent good solutions for everyday room temperatures of 19– 22 hC and interior humidities of 50 – 60 % (i.e. in office and residential buildings, including kitchens, bathrooms, etc.) and at the same time comply with the minimum thermal resistance requirements with respect to vapour diffusion. A vapour barrier is unnecessary with such forms of construction. The building physics requirements of buildings with specific requirements (cold stores, swimming pools, etc.) must be given special consideration. The diffusion behaviour of the wall construction must be checked for the wintertime thermal performance (Fig. 2.100). By contrast, concrete walls with internal thermal insulation (Fig. 2.99c) and a lining of plasterboard are generally insufficient because condensation collecting on the cold internal surface is excessive and therefore cannot dry out properly. A vapour barrier (e.g. aluminium foil) on the inside of the thermal insulation, i.e. between insulation and lining, is essential in such situations. Such a vapour barrier may also be necessary in conjunction with a thick facing leaf. The vapour diffusion can be improved by employing a façade with a ventilation cavity, i.e. an air space between facing leaf and thermal insulation (Fig. 2.99e), instead of a sandwich construction. This cavity, which should be at least 4 cm wide, allows water vapour to escape to the outside air. This form of construction permits the use of a much denser facing leaf, e.g. ceramic tiles, even sheet metal.

Fig. 2.99

Types of façade configuration

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Fig. 2.100 Temperature and pressure gradients plus the zones of usable thermal mass depending on the position of the layer of thermal insulation [116]

If, for example, for production reasons, sheeting is required adjacent to the thermal insulation in order to contain, for example, mineral fibre insulation, then this may only be placed on the warm side of the layer of thermal insulation. Whether the more expensive mineral fibre insulation needs to be used for the thermal insulation at all is another matter, however. Polystyrene is less expensive, easy to work and also unaffected by water, but it is combustible. An incombustible material is therefore required around windows to prevent the spread of fire. As a rule, edges must be finished with an incombustible thermal insulation material or a fire stop detail. The acoustic behaviour of concrete façades is essentially governed by the windows and not by the precast concrete elements themselves. This aspect will not be looked at further here. 2.4.2

Façade design

Apart from performing the building physics functions, a façade must also frame the windows. Fig. 2.101 shows basic forms for the segmentation of a façade for windows and joints. The simple horizontal ribbon façade can be varied in different ways (Fig. 2.101, left). The fenestrate façade is the typical form for large-panel construction. However, the loadbearing façade extending the full height of the building is seldom encountered in Germany, in contrast to the USA, which means that German architects have not yet fully explored the possibilities of this type of façade (Fig. 2.102). Façade panels can be supported on the edges of floor slabs, or in the form of an L-shaped loadbearing panel can themselves span from column to column and support the floor slabs, or in the form of a loadbearing wall with internal corbels form the loadbearing structure of the external wall. The façade shown in Fig. 2.103a is a horizontal ribbon type with window mullions but no joint intersections. Here, the continuous loadbearing columns at the same time serve as façade design elements and the spandrel panels span-

105


Fig. 2.101

Segmentation of façades for windows and joints

Fig. 2.102

Loadbearing façades [119]

ning between these are in the form of loadbearing L-beams with thermal insulation attached at the precasting plant. The facing leaf to the spandrel panels was mounted later (see also Figs 2.116 and 2.125). Joint intersections are likewise absent from the façade shown in Fig. 2.103b. Fig. 2.103c shows a horizontal ribbon façade that incorporates external fire escape balconies. In the USA, loadbearing façades that extend the full height of the building are especially popular for buildings up to three storeys high, likewise wide, storey-high loadbearing panels for multi-storey buildings. The reinforcement required just for demoulding, transport and erection is in most cases perfectly adequate for loadbearing purposes. Production, transport and erection limitations restrict multi-storey wall panels to a height of about 12–14 m. Buildings with loadbearing façades up to 20 storeys high have been built in the USA [119].

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(a) Zu¨blin-House, Stuttgart (architect: Bo¨hm)

(b) Publicitas office building, St. Gallen (architects: Danzeisen/Voser/Forrer)

(c) DYWIDAG office building, Munich (architects: Ba¨tge & Stahlmecke)

Fig. 2.103

Various façade configurations


107

Two-storey-high loadbearing façade panels were used for a 12-storey hospital in Chicago, with adjacent panels offset by one storey in order to avoid continuous horizontal joints. Loadbearing façades are particularly economic when they can also provide building stability functions and strengthening ribs fit into the architectural concept. However, the thermal insulation to a loadbearing façade usually has to be attached on the inside, with the associated disadvantages. It is always more sensible to position the loadbearing structure within the thermal insulation of the building envelope (see Fig. 2.104 and [120]). It should be remembered here that the windows must be attached to the loadbearing structure, or the loadbearing leaf of the sandwich panel, and not to the facing leaf, which is subjected to deformations. Only in the case of fenestrate façades can windows also be attached to the facing leaf. Waterproof seals and avoiding thermal bridges around windows are aspects that must be given due attention. One area that is still severely neglected is the design of precast concrete façades with respect to their weathering and ageing behaviour. Many sins were committed in the past which have given precast concrete construction a poor reputation. This can certainly be attributed to German architects’ poor acceptance of construction with concrete façades, in contrast to their colleagues in the USA, for instance. This leads to the numerous design options not being recognised and therefore not being included in university curricula. It is obvious that, as with stone façades, we cannot prevent the weather from having an effect on the surface. And the weathering effects are similar to those of stone or brick buildings.

Fig. 2.104 Sandwich panels: position of thermal insulation at columns

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Fig. 2.105 Uniform soiling remains acceptable; it underscores the structure of the façade or coincides with the existing shadows [121]

However, the accumulation of dust and dirt is normally less visible on the small-format patterns of bricks or stones than on the much larger and often smooth surfaces of concrete façades [119]. Therefore, the aim should be to achieve a “dignified ageing” – a skill of the masterbuilders of the Middle Ages – through the choice of the right façade structure, segmentation and detailing. Soiling is unavoidable and regular cleaning is expensive. So we should be aiming at achieving uniform soiling that possibly and hopefully underscores the structure of the façade. We then speak of a patina. To achieve this, we must consider the potential water run-off for the façade (Fig. 2.105) [121]. Water run-off is almost always the sole cause of objectionable soiling. The water must therefore always be channelled or “camouflaged” by the structuring of the surface. Water should never be allowed to remain on concrete surfaces; adequate falls must always be included so that the water can drain away. The quantity of rainwater, its velocity and its angle of incidence are different on each side of a building, and also vary over the height of the building. So, as with other structures, we cannot expect all parts of the building to exhibit the same degree of ageing. When considering the construction details, we must pay special attention to sloping surfaces, projections, rainwater drips, parapets, eaves and verges, also surface textures and colours, window openings and joints. Glass in particular should be protected against water draining from concrete surfaces. The ensuing hydroxides (alkalis with high pH value) can etch glass in air. Rainwater drips according to Fig. 2.106 must be incorporated into facing leaves at horizontal or low-pitch window head details. Façades with a sufficiently coarse surface texture and glass surfaces set back deep within the façade structure generally exhibit uniform soiling. Flat roofs require parapets with a minimum overhang to prevent wind driving rainwater over the roof and onto the façade. The top edge of a parapet must fall back towards the roof and must be finished with a sheet metal capping that projects min. 15 mm beyond the façade to form a rainwater drip. The joints in the sheet metal capping should always

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Fig. 2.106

Rainwater drip detail at window head

coincide with the real or dummy joints in the parapet elements in order to avoid ugly streaks on the concrete façade. The surface finish is important for the ageing behaviour as well as the initial appearance. Smooth concrete surfaces are harsh and unsightly, and quickly develop streaks during rainfall. Although more dust and dirt can collect on exposed aggregate surfaces, their appearance nevertheless remains acceptable. The grains of aggregate interrupt and distribute the run-off water and therefore prevent unattractive streaks. Vertical ribs in the surface structure help to ensure a controlled, vertical water run-off and prevent uncontrolled, sideways distribution. Dust and dirt collect in the grooves between the ribs and thus emphasize the ribbed structure (Figs 2.107 to 2.109). All these surface designs should of course consider the requirements of production in order to guarantee an economic façade.

Fig. 2.107

Office building in Munich (Hinteregger)

110


Fig. 2.108

Office building in Munich (Held + Franke)

Fig. 2.109

Office building in Rijswijk, Netherlands (HIBE)


111

Fig. 2.110 University of Riyadh, Saudi Arabia (a) Erecting a façade panel (b) Institute building (c) Casting a façade panel in a rigid mould

Here again, the maximum permissible transport dimensions and erection weights should be exploited to the full. The smaller the elements, the greater their number, which in turn results in more operations during loading/unloading and erection, more fixing points and more joints, and hence higher costs. If the architecture requires the scale of large façade elements to be reduced, then the answer is to include dummy joints (Fig. 2.110). Economic production is achieved by being able to remove façade elements from rigid moulds, which means a min. 1:10 taper on all edges and openings. This taper should be increased to min. 1:5 in the case of several openings per element or ribbed panels. All arrises in contact with the mould should be chamfered. Likewise, the transition from rib to main body of panel should be rounded if at all possible in order to prevent cracking (Fig. 2.111).

112


Fig. 2.111

2.4.3

Façade element details

Joint design

The joints between the façade elements represent an intrinsic part of the entire building envelope (see also [122–125] and FDB leaflet No. 3 on the design of precast concrete façades – available in German only). The joints are the weakest link with respect to the waterproofing and airtightness of the whole wall. Their simple design and construction is therefore critical for production and erection. Joint widths should not be chosen merely from the point of view of their appearance, but rather designed to suit the size of the element, the manufacturing tolerances, the jointing materials and the flanks of the joints. It is not advisable to reduce the sizes of elements in order to reduce movements at joints. On the contrary, it is better to plan for as few joints as possible because this approach is certainly more economic, also with respect to the cost of maintenance. Joint waterproofing must satisfy the following requirements [122]: x

x

x x

x x

The joint detail must be able to accommodate all movements resulting from temperature and moisture fluctuations, possibly also settlement, without damage. The joint waterproofing must satisfy the building physics requirements with respect to thermal insulation, sound insulation, moisture control and fire protection (DIN 4108, DIN 4109, DIN 4102). The joints must be able to compensate for production and erection tolerances. It must be possible to install the joint waterproofing irrespective of the weather conditions. The joint waterproofing must be permanent. The joint must satisfy architectural and economic demands.

Movements due to temperature and moisture fluctuations amounting to approx. 1 mm/m wall length can be expected for concrete façades. We generally distinguish between four methods of waterproofing the joints in concrete façades:


113

Fig. 2.112 Recommended values for designing the joint width and permissible minimum joint widths for buildings according to DIN 18540 Table 2

a) Joint waterproofing with elastic sealing compounds (e.g. Thiokol) to DIN 18540 (Fig. 2.112)

Sizing joint widths for joints with sealing compounds is carried out taking into account the fact that the sealing compounds may not be overstretched, i.e. Db/b I 25 %. The table in Fig. 2.112 lists the nominal values for planning purposes and the minimum joint widths in the finished structure according to DIN 18540. Such joints can be used virtually anywhere and do not place any particular demands on the wall construction. However, they are sensitive to large tolerances, they can only be installed at certain outside temperatures (5 hC I T I 40 hC, dry wall edges), which makes them unsuitable for the countries of the Middle East, for example, and their durability is limited. Fig. 2.113 shows joints between sandwich panels. The horizontal joint in the loadbearing leaf of the panels is filled with cement mortar. The outer sealing compound is installed on a closed-pore foam strip which is inserted first. b) Drained joints

In this type of joint the sealing function is essentially achieved by the shape of the wall panel edges. The horizontal joint is in the form of a “threshold” that is high enough to act as a barrier against driving rain, i.e. preventing wind forcing the rain beyond the higher part of the joint. This results in a minimum threshold height of 10 cm for the highest moisture load group to DIN 4108-3 (which applies in coastal regions and the foothills

114

Fig. 2.113 Joints in sandwich panel walls filled with permanently elastic sealing compound


Fig. 2.114 Drained joint: with step in horizontal joint and sealing strip inserted into vertical joint afterwards

of the Alps, also to high-rise buildings), in all other cases min. 8 cm. The joint should be 1.0 –1.5 cm wide and the angle of the front face of the threshold should be i 60h, preferably 90h. In addition, the joint should be made windproof by inserting a mineral wool rope or mortar packing between the wall panels. The vertical joint is in the form of a pressure-equalising (i.e. ventilated) joint, as shown in Fig. 2.114. PVC channels are cast into the concrete wall panels into which a baffle is inserted during erection to form a barrier against the rain. The cast-in channels serve simultaneously as pressure equalisation spaces in which rainwater can collect and drain downwards, escaping safely to the outside at the next horizontal joint. A wind barrier according to Fig. 2.114 is necessary where the joint passes through the entire wall construction, also through the loadbearing leaf. Drained joints of the type described above are generally unaffected by tolerances and unforeseen joint movements as a result of settlement or even earthquakes. Such joints can be completed irrespective of the weather conditions. Baffles are durable and can also be obtained in any RAL colour provided the order is large enough. Fig. 2.116 shows quite clearly how a prudent architectural design with half-round strengthening sections at the edges of the elements create space for the inclusion of a groove into which a baffle can be inserted [126].

115


Fig. 2.115 Joint sealed by bonding loop of elastomeric sealing material to both sides of joint

c) Waterproofing with adhesive strips (Fig. 2.115)

Recent years have seen the development of joint waterproofing in which the joint is covered with elastomeric strips made from polysulphide, polyurethane or silicone. Firstly, a sealing compound made from the same material as the sealing strip is sprayed on the flanks of the joint. The sealing strips are subsequently pressed into the compound, preferably forming the strip material into a slight loop while doing so. The advantage of forming a loop is that movements of the joint flanks as a result of temperature fluctuations do not subject either the sealing strip or the adhesive compound to tension or shear. With jointing materials chosen to match the colour of the rest of the façade, this type of joint also complies with aesthetic demands.

Fig. 2.116

Façade intersection sealed with preformed profiles on Zu¨blin House

116


Fig. 2.117 Joint sealed with precompressed sealing strip (illmod system)

d) Joint waterproofing with precompressed sealing strips (DIN 18542:2009) (Fig. 2.117)

This type of waterproofing involves bonding a precompressed sealing strip made from impregnated polyurethane foam to one side of a joint before erecting the next panel, or inserting it into the finished joint afterwards. The precompression is subsequently relieved and the strip seals the joint, within given tolerances. The effect is therefore not chemical, but rather purely physical. Relieving the precompression takes place faster in warm weather and correspondingly slower in cold weather. 2.4.4

Façade fixings

When it comes to façade fixings ([127] and [128], FDB leaflet No. 4 covering fixing methods for precast concrete façades – available in German only) we make a distinction between the following two functions: – Retaining anchors as ties between the inner and outer leaves of sandwich panels or façades with cavities – Fixings as connectors between the façade elements and the building’s loadbearing structure The materials for these two types of fixing must comply with very high specifications (see also FDB leaflet No. 2 covering corrosion protection of inaccessible steel connectors for precast concrete elements – available in German only) because façades with defective fixings represent a considerable risk to the public. The majority of the parts of façade fixings are inaccessible for maintenance and repairs after erecting the faç ade, at best only with great effort and at great cost. As already mentioned, all components used in the building envelope are directly exposed to the weather and temperature changes. Those are the reasons for the main demands placed on façade fixings: they must be made from a permanently corrosion-resistant material and designed in such a way that they can accommodate temperature-induced structural movements without fatigue.

117


Stainless steel with special properties is compulsory for façade fixings outside of the loadbearing leaf with its insulating and waterproofing functions. Only steel grades 1.4401 and 1.4571 to DIN EN ISO 10088 und DIN EN ISO 3506 may be used (see also DIBt Approval Z 30.3- 6 covering fasteners and components made from stainless steels). Stainless steels with the well-known manufacturer’s designation V2A do not comply with this requirement – only V4A steels. Type-tested fixings always consist of steels “approved for anchorages in reinforced concrete construction”. Always adhere to the information given in the respective approvals when designing and machining stainless steels because these sometimes differ considerably from normal structural steels. 2.4.4.1

Retaining anchors for sandwich façade panels

The function of the retaining anchors is to tie together the three layers of the sandwich panel and in doing so carry all the forces that occur. These forces occur as a result of the selfweight of the panel (which can act differently in every position from demoulding to erection), changes in length and deformations due to temperature changes, and finally the pressure and suction effects of the wind. Fig. 2.118 shows the basic layout that should be used for retaining anchors. According to this, one support anchor is positioned as close as possible to the panel’s centre of gravity and the nail-type retaining anchors are distributed over the rest of the area, acting as spacers and accommodating deformations through their elastic bending capacity. A huge range of retaining anchor systems are available on the market (Fig. 2.119), for which the results of type testing are generally available. The self-weight of the facing leaf is applied to the loadbearing leaf with an eccentricity – and where there is a ventilation cavity the eccentricity is increased by the width of that cavity (generally 4 cm). With certain fixing systems it is important to take into account the fact that when demoulding the façade panel, the self-weight can act on the fixings at an angle that is 90h to that of the final condition and they may have to carry additional adhesive stresses. Special fixings to cope with this may be necessary in some cases. Eccentricities within the plane

Fig. 2.118 Schematic layout for retaining anchors [122]

Fig. 2.119 Support anchor principles of various manufacturers

118

Fig. 2.120


Corner details for sandwich panels

of the facing leaf should be avoided wherever possible. Torsion anchors are generally required where unintentional eccentricities due to inaccuracies, openings or erection loads must be accommodated. Wind loads on type-tested retaining anchors are generally based on DIN 1055- 4; the new 2005 edition with its higher suction coefficients for the corners of buildings should be consulted. Special investigations are usually required when facing leaves cantilever beyond the loadbearing leaf. Uniform temperature deformations cause bending moments in the retaining anchors. According to DIN 18515 “Cladding for external walls”, a temperature difference of e50 K should normally be allowed for. Where a facing leaf passes around a corner of the building, then a suitable gap should be left during production to permit the facing leaf to deform without restraint (Fig. 2.120). Such a gap is unnecessary with short corner cladding panels whose movement fulcrum is at the corner. The temperature gradient DT through the thickness of the facing leaf, which can occur several times each day, causes curvature (Fig. 2.121) and consequently tensile or compressive forces in the retaining anchors which increase with the thickness of the facing leaf. Facing leaves should therefore be no thicker than 8–10 cm, with the larger figure applying to profiled façades in exposed aggregate concrete. The temperature gradient is greater in façades with a ventilation cavity because there is no helpful build-up of heat in front of the layer of insulation as is the case with sandwich panels. Additional calculations will be necessary if it is not possible to achieve a minimum-restraint anchorage with a four-point retaining anchor system [129–131].

119


Fig. 2.121 Deformation of unrestrained facing leaf as a result of DT [122]

The reinforcement in the facing leaf generally consists of one layer of minimum reinforcement, but additional reinforcement is generally necessary in the vicinity of the support anchor. Additional reinforcing bars are recommended around the perimeter and around window openings in order to control cracking [122]. Including an additional bar at 45h at each window corner to control cracking is usually not possible because a concrete cover of 3.5 cm to the outside should be regarded as the minimum. The retaining anchors, generally round steel dowels, should be distributed as evenly as possible over the surface, preferably on a square grid. Additional anchors along the edges may be necessary to cope with demoulding. However, providing more anchors than is absolutely necessary should be avoided because they have a negative effect on the thermal resistance [132]. A suitable concrete mix and a low-shrink concrete are important considerations for the production of façade panels. Adequate curing is important for any facing leaf that projects beyond the loadbearing leaf and is therefore exposed on both sides (see section 4.3.1). Facing leaves with smooth surfaces should be limited to a length of 5– 6 m, indeed 3.50 m is recommended in [120]. Somewhat longer lengths are sometimes possible with heavily structured surfaces where smaller, less visible cracks are acceptable or defined cracking points are created by including dummy joints. Facing leaves suspended free from restraint in front of a ventilation cavity can be built with longer lengths without joints than is the case for the facing leaves of sandwich panels. A ventilation cavity can be created with the help of special 40 mm thick studded sheets or by installing polystyrene blocks (approx. 4 pcs./m2). Timber wedges, which are removed again after demoulding, are used where larger quantities are being produced (Fig. 2.122) [127]. Ref. [133] the use of sand fillings to create an air space. 2.4.4.2

Fixing façade panels

The fixings for the loadbearing leaves of façade and spandrel panels must be designed for the dead loads, possibly with surcharges in seismic zones, and for wind pressure and suction. Special attention must be paid to forces due to constrained shrinkage and, possibly, friction forces due to disparate movements between façade and structure. Façade panels are either supported from below or suspended from above (Figs 2.123 and 2.126). In both cases they then only require to be held at the sides. The overturning moment must be taken into account in the case of an eccentric support, e.g. on corbels or boots. When panels are supported from below, the loads of any panels above may also need to be considered, whereas when suspended from above, the reinforcement must be

120

Fig. 2.122


Precasting a facing leaf with ventilation cavity (Frimeda)

designed to carry self-weight at least. In doing so, however, the cracking load of the concrete should never be exceeded. Fixings must be able to accommodate erection tolerances of min. e2.5 cm. They should be designed in such a way that erection ties up the crane for only a short time and the final alignment and tightening can be carried out once the panel has been detached from the crane hook. Erection personnel should not need any special scaffolding and should not


121

Fig. 2.123 Façade fixings: (a) Façade suspended from above (b) Loadbearing leaf supported from below

be exposed to any unnecessary risks during the positioning and fixing. Fasteners with adequate protection against corrosion are essential (see FDB leaflet No. 2 covering corrosion protection of inaccessible steel connectors for precast concrete elements – available in German only). A multitude of fixings for façade panels is available on the market. It is possible to distinguish between certain basic types depending on their configuration and method of carrying the forces. However, each type has its own particular merits and demerits [133]: 1. Connections using cast-in reinforcement, where the façade is connected monolithically to the floor slab (Fig. 2.124).

Advantages: large tolerance adjustment options, good protection against corrosion, good fire resistance, economical production. Disadvantages: a temporary retaining fixing is required during erection which can in turn have a serious negative effect on the overall economy. 2. Welded connections

Advantages: very easy to adjust on site. Disadvantages: no chance for façade to deform with respect to loadbearing structure, risk of cracking in the vicinity of the welds, trained welders needed on the building site at the right time, problem of designing the connection to achieve an adequate fire resistance, ties up the crane for a long time during erection. 3. Boots and corbels Continuous boots (Fig. 2.123a) or individual corbels (Fig. 2.125). Special shims are necessary when the expected movements are large. Fixing is achieved mostly by way of cast-in dowels that are inserted through the boot/corbel and subsequently

122


Fig. 2.124 Façade held in place with cast-in reinforcement: (a) offices and retail building, Stuttgart (contractor: Zu¨blin), (b) detail of façade fixing with cast-in reinforcement

grouted; a plastic sleeve can help to achieve a certain degree of horizontal deformation. Boots/corbels can be positioned at the top or the bottom. They can also be in the form of cast-in steel sections.

Advantages: easy and quick form of connection, possibility of later adjustability without the need to hire expensive cranes, good corrosion protection and adequate fire resistance, limited tolerance accommodation, but this can be doubled if merely a hole for

Fig. 2.125 Façade panels suspended from separate corbels (Zu¨blin House) (1) Façade column (2) Internal column (3) Spandrel panel facing leaf (4) L-beam with thermal insulation (5) Thermal insulation to column (6) Inverted channel section floor unit (7) Precast concrete plank (8) Preformed joint sealing strip (9) In situ concrete


123

Fig. 2.126 Façades supported on column corbels (Fachvereinigung Deutscher Betonfertigteilbau e.V.)

the dowel is initially left in the floor slab construction and the dowel subsequently inserted through the holes in both panel and slab and then grouted in place.

Disadvantages: creation of a thermal bridge at the boot/corbel, certain difficulties in ensuring adequate deformation possibilities. 4. Suspension anchors The above connections all have one disadvantage in common: it is difficult for them to accommodate shrinkage and temperature changes without causing restraint forces. To overcome this, they are therefore fitted only in loadbearing leaves in thermally insulated parts of the structure. Zero-restraint support for a facing leaf in front of a ventilation cavity can only be achieved with articulated suspension anchors (Fig. 2.127).

Advantages: small thermal bridge, flat façade panels with no protruding corbels/boots are better for storage and transport, easy to adjust for tolerances, good for suspended façades with ventilation cavities, can be subsequently secured with anchors, adjustable in all three directions. Disadvantages : relatively expensive forms of construction using stainless steel, difficult to obtain adequate fire protection.

124

Fig. 2.127 Façade panels fixed with suspension anchors: (a) University of Tu¨bingen (contractor: Zu¨blin), (b) façade details



2.4.5

Architectural façades

2.4.5.1

Decorative façades employing precast concrete elements

125

An artistic façade architecture using precast concrete elements has become established in recent years. Only in some cases do these façades perform a loadbearing function. In many cases it is the architecture of the building that is of primary importance. The diverse architectural options of concrete as a building material plus the advantages of factory production, e.g. excellent quality, are exploited in this development. The architectural and functional options are to be found in the – – – –

almost unlimited mouldability of concrete, the different colours possible with concrete, the load-carrying capacity of concrete, and the high weathering resistance of high-strength concrete.

The development of high-strength and self-compacting concretes create the opportunities for good durability and excellent surface finishes that can only be properly implemented in conjunction with factory production. Three up-to-date examples of such façades are shown below. The construction details are essentially in line with the aforementioned boundary conditions and are therefore not explored in detail. The aim here is only to outline the particular features.

Phaeno Science Centre, Wolfsburg [136] (Fig. 2.128) Large areas of the façade were constructed from precast concrete elements mounted on a structural steel frame. The main reason for the use of precast concrete elements was the demanding surface finish specification. Thermal insulation, a vapour barrier and a plasterboard lining were attached to the inside of each precast concrete element. Batch production, actually one of the great benefits of precast concrete elements, was irrelevant on this project. Each one of the 39 elements, which weigh up to 10 t, is unique.

Laboratory building, University of Wageningen [137] (Fig. 2.129) Unlike the Phaeno façade above, the honeycomb-like façade of this building is not only structural, but was also designed with the production requirements of precast concrete elements in mind. The standardised precast concrete components give the building a distinctive look and carry the loads of the floors via anchor plates and steel beams. The floor beams are thermally insulated to prevent thermal bridges. Besides the high repetition rate, the cross-sections taper towards the outside so that it was easier to remove them from the moulds without the need for dropside panels. The titanium dioxide mixed into the white, self-compacting concrete of grade B65 is intended to create a self-cleaning effect and thus avoid unsightly streaks and dirty edges. The expansion joints for the façade are located at the corners of the building.

126


Fig. 2.128 Erecting a cranked façade element (Phaeno, Wolfsburg; architect: Zaha M. Hadid)

Fig. 2.129 Honeycomb-type façade to laboratory building at University of Wageningen, Netherlands (architect: Rafael Vinoly)

Community centre in Mannheim-Neuhermsheim [138] (Fig. 2.130) Single-storey precast concrete elements form the façade to this community centre. They carry the roof loads and the wide gap between the concrete and the “thermal break” glass façade creates a covered walkway around the building. The junction between façade and roof is suitably thermally insulated (Fig. 2.130b). The precast concrete elements of the apparently random façade were actually produced using just two basic moulds. The irregular arrangement was created by turning some elements upside down and employing an irregular spacing. 2.4.5.2

Façade panels made from high-strength and glass fibre-reinforced concrete

More recent developments concerning thin façade panels can be broken down into – façade panels made from ultra-high performance concrete, and – façade panels made from textile-reinforced high-strength fine-grained concrete. Actually, façade panels made from ultra-high performance concrete should be classed as ultra-high-performance concrete (UHPC) panels because it is not so much their high compressive strength that makes them worthwhile, but rather their excellent durability with respect to environmental influences. This material is not yet approved for general use in


127

Fig. 2.130 Precast concrete façade to community centre in Mannheim-Neuhermsheim (architects: netzwerkarchitekten; precast concrete contractor: Hering Bau)

Germany and can only be used in conjunction with an Individual Approval for a particular project. A number of projects have already been completed in France, however: faç ade panels with an area of up to 4.40 m2 and a thickness of just 20 mm were used for the Rhodia company’s development centre in Aubervilliers (Fig. 2.131a), and the façade of the RATP bus depot (Fig. 2.131b) has a surface finish reminiscent of LEGO bricks! The elements of this latter façade are 30 mm thick.

Façade panels with glass fibre textile reinforcement and made from high-strength finegrained concrete evolved out of research into building with textile-reinforced concrete (TRC). Developed in the 1990s, the principles for its loadbearing behaviour and design are now available, which means that further applications are now possible. Ref. [139] provides an insight into the technology of textile-reinforced façades (Fig. 2.134). The textiles are produced from engineered high-performance fibres and materials such as alkali-resistant (AR) glass, carbon or plastics. The individual fibres are called filaments and have diameters from about 10 to 30 mm. During production, hundreds or thousands of these filaments are bundled together to form so-called filament yarns. Engineering tex-

128


(a)

Fig. 2.131 Façade panels made from ultra-high performance concrete: (a) Rhodia development centre, Aubervilliers, France; architect: JF Denner; (b) RATP bus depot, Thiasis, France; architect: ECDM Emmanuel Combard Dominique Marree (b)

Fig. 2.132 Two-dimensional reinforcement as knitted fabric [139]

tiles, in the form of knitted or woven fabrics, are then manufactured from these yarns (Fig. 2.132). The production costs are still higher than those of conventionally reinforced concrete but the AR glass fibres represent the most economic alternative in this respect. The main advantage of TRC façade panels is certainly their thinness, brought about by the fact that there is no steel reinforcement requiring a certain concrete cover in order to guarantee protection against corrosion. The panels, which can be produced in thicknesses between 15 and 30 mm, are very low in weight, which in turn leads to savings in fixings. In particular, the thick layers of thermal insulation called for these days place substantial loads on anchor systems because of the greater eccentricity of the facing leaves or suspended façade panels. The fine-grained concrete mixes used, with maximum aggregate sizes of 1–2 mm, enable not only the production of high-quality fair-face concrete surfaces, but also keen-edged


129

Fig. 2.133 Sandwich element (during concreting and as finished component) [139]

Fig. 2.134 Façade made from betoShellr panels from Hering Bau [139]

Fig. 2.135 Front view of pebble-shaped venue for events (contractors: Schmid, Baltringen, and Rudolph, Weiler-Simmerberg)

parts and profiles that offer architects great design freedoms. The well-known surface finishes and colours possible with conventional concrete can continue to be used. The relatively high rate of shrinkage must be taken into account for larger components and narrow joint widths. It is possible to use TRC façade panels for both the facing leaves of sandwich elements and for suspended façade panels. Both types have already been used in conjunction with an Individual Approval on a building for RWTH Aachen University (Fig. 2.133). A National Technical Approval has already been granted for TRC suspended façade panels [140]. One critical factor is certainly the fixing of these very thin concrete elements – just 20 mm! Special cast-in anchors are used [141]. These panels can be produced with great accuracy up to sizes of 120 x 60 cm and require only very narrow joint widths of approx. 3 mm. Ref. [142] describes an unusual use of textile-reinforced façade panels. In contrast to conventional batch production, the outer envelope of a venue for events in Friedrichshafen was designed to imitate a pebble (Fig. 2.135). Accordingly, a total of 124 elements in var-

130


ious shades of anthracite and sizes up to 4.0 x 5.30 m were produced with a leaf thickness of 25 mm. A full-scale model made from polystyrene foam served as the mould. These parts, also manufactured and installed according to an Individual Approval, are connected to the structural steelwork frame by means of cast-in steel parts. The above examples show that research into new concretes and composite materials are showing promising developments for façade construction which present an alternative to steel-and-glass façades. The numerous architectural options are opening up a wide range of creative opportunities for architects to continue designing new types of concrete façade. 2.5

Connections

It is necessary to design connections so that the individual floor units, beams, columns and wall or façade panels can be assembled to form the loadbearing structure. In doing so, it is always vital to consider the needs of production and erection as well as the structural and constructional aspects. Apart from the architectural and building physics requirements, all connections must also consider the routing of building services. In a frame structure (Fig. 2.137) the beam/internal column connection is joined by the beam/perimeter column detail, and here the beams may be parallel with or at 90h to the façade. And for the façade itself, it is also necessary to clarify the junctions with the internal and external corner columns (Fig. 2.136). Such connections can have a considerable effect on the entire frame system, especially when the interior fitting-out grid is offset from the structural grid or when there are maintenance or fire escape balconies in front of the façade (Fig. 2.138). The aforementioned connections must be designed for the standard floor, raised ground floor and roof. Interior fitting-out details specific to buildings and single-storey sheds made from precast concrete elements can be found in [8]. Generally, only through teamwork is it possible to consider the diverse boundary conditions so that an optimum design can be achieved with optimum economy. Such cooperation between architect, engineer, building services consultant, precast concrete manufacturer and transport and erection contractors should take place as early as possible. A number of examples of connections are shown on the following pages, some of which are also shown in [6] (see also Fig. 2.148). For example, Fig. 2.139 shows a beam/internal column connection with corbels on just two sides and various beam types. As downstand beams are almost exclusively designed as single-span members with statically determinate support conditions, they require a correspondingly wide compression zone. This advantage is offered by a beam in the form of an inverted channel section. Another advantage of such a beam is that its webs can pass either side of the column and thus provide a zone for routing services (see Fig. 2.141). Furthermore, it can also cantilever beyond the final column without the need for corbels on three sides of the column (see Fig. 2.147). Inverted channel section beams with web ends that pass either side of the columns cannot be lowered into position from above (with the beam horizontal) in the case

2.5 Connections

Fig. 2.136

131

Extract from a catalogue for an industrialised building system (6M system, Zu¨blin) [97]

132


Fig. 2.137

Connections in a frame system

Fig. 2.139

Internal column/beam connection details

Fig. 2.138 Building corner detail with fire escape balconies

of multi-storey columns with corbels on the sides and so it is advisable to cut back the flange to such an extent (Figs 2.139b and d) that the beam can be positioned horizontally but at an angle between the columns and then twisted into position. If this is not possible, then lowering from above with the beam hanging at an angle (i.e. with diagonal pull) will be unavoidable. Rectangular beams should be connected to the floor slab, e.g. composite plank floors, via starter bars to form T-beams wherever possible. A considerable structural depth will be required for the floor where there are many services below the soffit, e.g. ventilation ducts in air-conditioned buildings. The most economic solution in such a case is to support floor slabs and beams without notches (Fig. 2.140a). However, in buildings with only a few building services below the floor, notching the floor slabs and ends of beams is usually preferred in order to minimise the structural depth of the slab and hence the overall height of the storey (Fig. 2.140b). For instance, double-T floor units can be notched over the full depth of their webs. Moreover, the web ends of double-T or inverted channel section floor units can be notched at an angle to provide some space for routing services (Fig. 2.142). The reinforcement in such a

133

2.5 Connections

a) Building with extensive services

b) Building with few services

Fig. 2.140

Floor designs and building services

134


Fig. 2.143 Junctions between double-T floor units and beams

Fig. 2.141 Void for building services in longitudinal direction of building

Fig. 2.142 Angled notches at ends of beams for accommodating building services

Fig. 2.144 Holes drilled through the 10 cm thick flange of a double-T floor unit after erection

case then continues upwards at an angle, following the flow of forces, and must be anchored accordingly (see section 2.6.2). Fig. 2.143 shows potential connections between double-T floor units and various beams. The following opening widths are necessary for services: Gas 5 cm Electrics 5–7 cm Water 7.5 cm Heating 15 cm Drainage 25–50 cm (1:50 to 1:100 fall) Ventilation ducts 0.50 –1.50 m2 (40 – 60 cm deep) Fig. 2.145 illustrates the principles for positioning services with respect to the loadbearing structure. Openings in the floor slab, where they are not standard openings adjacent to columns, for example, can be formed in the finished structural carcass by means of core drilling on site (Fig. 2.144). Normally, such drilling should not be carried out before the positions are approved in writing by the structural engineer. Forming the openings at the precasting plant

2.5 Connections

Fig. 2.145 members

135

Potential solutions for positioning building services in and adjacent to loadbearing

means that the large batches of floor units, or sometimes even beams, are disrupted. This is less a problem for production, where it is easy to form such openings, but more a problem for the organisation and the technical operations. On the general arrangement drawings such elements are only indicated by a different number; the openings are only shown on the individual element drawings. In addition, elements with various openings must be produced, stored and supplied at the same time as erection is taking place, which leads to considerable organisational input. Suspended ceilings are not required in housing and often not in office buildings with naturally ventilated individual offices either. In such cases flat soffits are necessary, which can be achieved with composite plank floors or hollow-core slabs. Junction boxes for ceiling lights can be incorporated in the planks at the precasting works, although this means that the connecting cables must be laid at the same time as placing the reinforcement for the concrete topping. Fig. 2.146 shows the connections between floor systems of conventionally reinforced hollow-core slabs and downstand beams or walls. The voids in the slabs can also be used for routing building services, and penetrations in the region of a void up to a width or diameter of 15 cm are possible without additional structural measures. Fig. 2.147 shows examples of different façade arrangements to accommodate heating systems. The connection details for Zu¨blin House are shown in Fig. 2.149. In this building the façade is structural, i.e. the columns are simultaneously loadbearing and decorative elements. The composite plank floors are supported on L-shaped perimeter beams from which the decorative façade panels were suspended in a separate operation. With this type of design, the building physics requirements become a critical design factor for the façade. The thermal insulation in the spandrel panels was attached to the outside of

136


Fig. 2.146 Connection details for a hollow-core slab

Fig. 2.147

Design examples for façade junctions (6M system, Zu¨blin)

2.5 Connections

Fig. 2.148

Connections (Fachvereinigung Deutscher Betonfertigteilbau e.V.)

137

138

Fig. 2.149


Construction principles for Zu¨blin House [44]

each perimeter L-beam at the precasting plant, whereas it is positioned internally at the columns (with suitable vapour barrier, see also Fig. 2.104) and was installed afterwards on site. The remaining thermal bridge at the column corbel is in this case less of a problem because this is an inward-facing corbel and does not result in a cooling rib effect as would be the case with, for example, an outward-facing beam boot. The composite plank floors in this building are supported on inverted channel section units in the middle bay, which above the corridors provide space for the building services (which are concealed behind a suspended ceiling). There are no suspended ceilings in the offices of this building. The junction boxes for the surface-mounted luminaires were cast in during production of the planks. All other electricity supplies are provided via conduits below the window sills and switches were incorporated in the lightweight corridor walls at the fitting-out stage. The constructional problems posed by connections are dealt with in more detail in the next section.

139

2.6 Current design issues

2.6

Current design issues

2.6.1

Additions to cross-sections, floors with concrete topping

Rectangular beams are in many cases subsequently combined with the floor slab to create T-beams (see, for example, Fig. 2.150). Likewise, prestressed double-T floor units are given an approx. 7 cm deep concrete topping in order to achieve a flat floor surface. DIN 1045-1:2001 [143] calls for a structural check to be carried out on all construction joints, so-called shear joints. This applies to the joints between precast concrete beams and in situ concrete as well as the planar joint between the floor units and the subsequent in situ concrete topping. But the various changes to the standards in recent years have led to uncertainties in practice [145]. The provisions given in DIN 1045-1:2001 and DIN 1045-1/A1:2008 will be outlined below in order to clarify the current design requirements. The analysis given in DIN 1045-1/A1 was altered substantially again in order to make it similar to the method in EC 2. For further information please refer to [195, 197]. All methods of verifying the shear joint are based on a total of three loadbearing components for transferring the shear action: – adhesion (bond) – friction (as a result of external axial stress) – reinforcement (shear-friction theory) (see also Fig. 3.30, p. 203) Moreover, there are further load-carrying mechanisms such as dowel action and the kinking effect (diagonal tension effect) which, however, are not included in the analysis. The adhesion component acts in the main before any cracks appear in the joint, whereas the reinforcement only starts to carry most of the load as the cracks widen. Therefore, DIN 1045-1:2001 assumes that only one of these two effects may be assumed in any particular case. Reinforcement is unnecessary when the shear force present does not exceed the following value: 1=3

vRd,ct w [0,042 h1 bct f ck s m sNd ] b

(20)

Here, only the actual joint between old and new concrete may be used for the joint width (see Fig. 2.151b). As this adhesion component does not normally accommodate the shear force present, especially for the shear action in the beam, the reinforcement required must be calculated from the following: vRd,ct w as fyd ðcot u S cot aÞ sin a S m sNd b

(21)

where the inclination of the strut should be assumed to be as follows: 1,0 J cot u J

1,2 m s 1,4scd =fcd 1svRd,ct =vEd

(22)

140


Fig. 2.150 T-beam slab made from precast concrete beams, precast concrete planks and in situ concrete topping

Fig. 2.151 Shear joint to DIN 1045-1: (a) with shear key (joggle or castellated joint), (b) examples of definition of joint width, (c) shear force diagram showing joint reinforcement required, (d) comparison of design results


141

The adhesion and friction coefficients plus the surface conditions should be taken from section 10.3.6 of DIN 1045-1. In order to take into account the difficult conditions regarding the quality of the joint design, especially in the case of smooth joints, DIN 1045-1 limits the angle of the strut to 45h. If u i 45h, the joint is not permissible and the design of the joint must be changed. Similarly to EC 2, the simultaneous effect of all loadbearing components is assumed when designing to DIN 1045-1/A1:2008 [145]. Removing the limit to the strut angle means that now even smooth joints can be realised with appropriate reinforcement. However, it should be pointed out that only a few test results are available for smooth joints (and even fewer for very smooth joints), which means that the appropriate care must be exercised with such joints [146]. The design value of the shear force acting at the interface between concrete topping and precast concrete element is calculated as follows: vEd w

Fcdj VEd Fcd z

where VEd Fcd w MEd/z Fcdj z

(23)

design shear force longitudinal force in flange cross-section under consideration longitudinal force component in additional cross-section lever arm for internal forces

The design value for the admissible shear force

vRdj w h1 cj fctd s m sNd b S vRdj,sy J vRdj, max

(24)

is made up of the following components: – Adhesion h1 · cj · fctd, with cj taken from Table 2.8 and the design value for the tensile strength of the concrete fctd w fctk,0.05/gc, where gc w 1.8 for unreinforced concrete; the adhesion component may not be included if the joint is subjected to tension or if dynamic loads are present. – Friction m · sNd, where m is the friction coefficient taken from Table 2.8 and sNd is an axial stress acting on the shear joint (pressure negative), although sNd values i 0.6 · fcd may not be used. – Reinforcement according to the so-called shear-friction theory (see also Fig. 3.30). This assumes that with a cracked joint and relative movement of the concrete parts, the reinforcement across the joint is loaded in tension because of the roughness of the joint. This gives rise to a compressive stress in the joint. Consequently, in principle the reinforcement fulfils the same function as an external compressive stress applied perpendicular to the joint surface. The dowel effect of the reinforcement is not considered here because it contributes relatively little to the shear capacity of such a joint.

142


Table 2.8 Coefficients for shear joint design to DIN 1045-1/A1 Line

Column

1

2

3

Surface characteristics to 10.3.6(1)

cj

m

n

1

shear key

0.50

0.9

0.70

2

rough

0.40

0.7

0.50

3

smooth

0.20

0.6

0.20

4

very smooth

0.00

0.5

0.00

vrdj,sy w ð1,2 m sin a S cos aÞ as fyd

(25)

where a angle of reinforcement crossing the joint

In order to avoid the failure of the inclined strut, the upper limit for the admissible shear stress is given by the following equation: vRdj, max w 0,5 n fcd b

(26)

However, this value, too, must be seen in the context of the roughness of the joint because otherwise every joint, no matter how smooth, could be loaded up to the effective value of the shear force provided adequate reinforcement was included. According to DIN 1045-1/A1 section 10.3.6, shear joints are classified according to the roughness of the contact surface as follows: x

x x

x

Very smooth – contact surface of precast concrete cast against steel or smooth timber mould. Smooth – surface trowelled or produced using slipforming or extrusion methods. Rough – surface roughened with a rake after concreting (3 mm tines at approx. 40 mm spacing), or by exposing the aggregate, or by other methods that result in an adequate loadbearing behaviour; see also DAfStb booklet 525 [147] (German only) for definitions of surface roughness. Shear key – although this must be in the form as shown in Fig. 2.151a, or by using aggregate of size dg j 16 mm and exposing it over a depth of at least 6 mm.

When casting an in situ concrete topping or concrete-filled joint, the surfaces must be free from cement laitance, sawdust, ice and oil. Dry surfaces should also be avoided. The consistency of the in situ concrete should be soft or fluid, and the concrete must be carefully compacted. Table 2.8 lists the coefficients for adhesion, roughness and strut failure. Fig. 2.151d compares the amount of shear reinforcement required and the maximum permissible shear force.


143

In principle, a shear connection to ensure composite action can be realised by providing steel reinforcement in the form of shear links, for example. However, special lattice beams are generally used for planar-type shear joints, which owing to their particular properties require a National Technical Approval (see also section 2.3). When designing a lattice beam to function as reinforcement for composite action and shear, it is important to take into account the fact that this is limited to the following shear force: VEd I 0,30 VRd, max

(27)

Reinforcement in the form of shear links should be provided for at least 50 % of the shear force in the case of greater loads. Lattice beams alone can function as reinforcement ensuring composite action. Therefore, there must be a distance of min. 2 cm between shear joint and top chord of lattice beam. 2.6.2

Corbels and notched beam ends

Corbels on columns or walls in conjunction with notched beam ends represent the most common type of connection in precast concrete frame construction. Column corbels, i.e. where the shear force is transferred directly downwards into the column via an inclined strut, are generally designed according to a diagonal strut model as shown in Fig. 2.152. Compared with a beam in bending, the corbel represents the special case of a very short cantilever. Experimental studies [156] have shown that it is possible to carry a much greater load than would be the case with a beam in bending. The reason for this is the diagonal strut leading directly into the supporting component, which is well restrained at the base of the corbel. With standard geometries and proper reinforcement arrangements, such a corbel will fail by first cracking at the internal corner between the corbel and the column above, with a subsequent reduction in area of the strut at the junction with the column below until it fails completely. The introduction of DIN 1045-1 in 2001 now means it is possible to design using linear member models. Fig. 2.153 shows the mechanical diagonal strut model and the design procedure. Examples can be found in [194] and [195]. The generalisation and simplification required for this type of model leads to a design that lies on the safe side. Actually, studies [156] have revealed that the compression zone at the bottom corner of the corbel reduces to a much greater degree than predicted by the model. Therefore, the design according to Fig. 2.153 leads to an increase in the amount of tension reinforcement required, especially in the case of stocky and heavily loaded corbels (Fig. 2.155). The experiments of Steinle [158] have shown that an inner lever arm of 0.95 · d is established upon failure as a result of the reduction in area of the compression zone. For the design it should therefore be sufficiently accurate to assume an inner lever arm z w 0.85 · d and carry out the design of the upper tie using the following equation:

144


Fig. 2.152 Flow of forces, diagonal strut model and typical reinforcement arrangements for corbels

Fig. 2.153

Design of corbels using the diagonal strut model [194] to DIN 1045-1:2001

145


Fig. 2.154 Determining the minimum corbel depth (according to [156])

T1 O

ac F S H i 0,5 F 0,85 d

If no sliding bearing has been specified, then a minimum horizontal force H w 0.20 · F should be assumed for the construction. The lower limit of 0.5 · F for the tensile force applies to deep wall corbels and in practice represents limiting the angle u of the strut to 60h with respect to the horizontal. Very deep corbels require main reinforcement not only near the top surface of the corbel. Assuming the compressive stress is limited to scw J 1.0 · fcd, the design of the diagonal struts in [156] leads to the definition of a minimum corbel depth as follows:

Fig. 2.155

Comparison of design approaches [156, 194–196]

146

min d j


3,58 FEd fcd bw

(28)

This value is not dependent on ac/h and is intended to apply as long as the design for bending of the cantilever does not require a greater depth. This limit lies at about ac/h w 1.1. When ac/h i 1.1, the design can be carried out according to Fig. 2.153, although where longer corbels are involved, there should be a direct transition to the design for bending of the cantilever. Far more important than determining the reinforcement exactly is, however, a properly engineered, sensible corbel detail. The designer should always remember that owing to the small dimensions of this component even small tolerances and deviations can in practice lead to considerable changes in the boundary conditions for the design. In principle, the utilisation should therefore be limited, the reinforcement should not be skimped and the planning and quality control work should be carried out carefully. The potential problems are totally disproportionate to the potential savings in reinforcement or concrete. The following design criteria must be taken into account: a) Specifying the corbel depth to limit the stresses b) Specifying the corbel length to ensure adequate anchorage of the tension reinforcement c) Combining the size of the bearing with the reinforcement layout d) Detailed planning of the reinforcement layout (scale drawings) The shear links in particular must be especially carefully planned. In principle, the shear links prevent the premature failure of the strut by resisting the tensile splitting forces. In the case of stocky corbels, horizontal shear links should be provided for a force of Fwd w 0,2 s 0,5 F

(29)

whereas it is the vertical shear links that become more and more critical as the slenderness of the corbel increases. Fig. 2.156 shows a reinforcement proposal depending on the slenderness of the corbel. The prerequisite for a well-functioning corbel is of course adequate anchorage for the tension reinforcement in the corbel. With large bar diameters this is best achieved by using a welded anchor plate (see section 3.2.1) or a welded transverse bar. For in situ concrete corbels it is recommended to specify a maximum concrete cover in addition to the usual minimum concrete cover in order to comply with tolerances for the position of the tension reinforcement. The following points are important if adequate anchorage is to be provided in the form of horizontal loops (the end of the incoming beam should be considered similarly in this situation and a bearing pressure of s j 0.2fck is assumed).


Fig. 2.156

147

Recommended shear link reinforcement for corbels (according to [195])

The loop (also known as a hairpin bar) should be overcompressed by the bearing pressure so that the usual concrete cover at 90h to the plane of the loop is adequate. In that case a loop diameter of dbr w 15ds (or a bending roller diameter dbr w 4ds in the case of link-type bars, i.e. with a straight section between the two bends) is possible for bar diameters ds J 16 mm (or dbr w 7ds for bar diameters ds j 20 mm). The reinforcement in the (if necessary) notched end of the beam should be anchored behind the front edge of the support. According to DIN 1045-1, the following applies: 2 2 As,erf lb,net w aa lb 3 3 As,vorh j 6ds for link-type shear reinforcement (dbr j 4ds) j 0.3 · aa · lb j 10ds for loops (dbr j 15ds) lb,dir w

(30)

where aa w 0.7 for link-type shear reinforcement (DIN 1045-1 Table 26) and aa w 0.5 for loops with dbr w 15ds and lb w

ds fyd 4 fbd

(31)

according to DIN1045-1. The shortest support length with lb,dir w 6ds is generally only possible when the reinforcement provided is more than two to three times that actually required. Similarly, the reinforcement in a column corbel must be anchored from the back edge of the support forwards to the end of the corbel over a length of lb,dir in the case of a constant bearing pressure. Consequently, this results in the minimum lengths for corbels and notched beam ends given in Fig. 2.157. Here, the maximum possible tolerance between beam and column must be taken into account for the dimension Dl. With sliding bearings the possible travel must

148


Fig. 2.157 Minimum corbel lengths for s j 0,2 fck

also be included. In addition, it should be ensured that the bearing pad is so soft that any unevenness in either contact surface is compensated for and the beam does not bear, for example, solely on the back edge (which could be the case with a prestressed cambered beam) or solely on the front edge of the support. In order to take into account non-uniform bearing pressures, the bends in both loops (beam and corbel) should begin at least on the axis of the bearing so that with the same loop diameter a complete circle is formed on plan (Fig. 2.157). Furthermore, the edge of the bearing should terminate a distance equal to the concrete cover dimension c1/2 back from the respective axis of the loop reinforcement: not much more, so that the loop remains overcompressed, and not less, so that the two opposing loops always overlap sufficiently. The minimum length of the corbel therefore depends on the required bearing length bL and the respective corbel or beam end clearances v and the maximum permissible tolerance Dl: l w bL S vTr S vK S max Dl

(32)


where vTr,K w

c1 ds c S S 2 2 2 Tr,K

149

(33)

If the anchorage length lb,dir is not adequate in this situation, the bearing length, and hence the corbel length, must be increased accordingly. It is essential to note that the construction tolerances permissible for the positioning of the reinforcement and the bearing pad and for the component dimensions tend to be used more or less to the full in practice. Therefore, where (small) dimensions are fully exploited, adherence to the dimensions should be guaranteed by quality control measures. Similarly, the corbel width is the total of the required bearing width tL plus the extra width on both sides c21 S d2s S c3 bw w tL S c1 S ds S 2c3

(34)

The reader should refer to section 2.6.5 for information on the size of corbels necessary to meet fire resistance requirements. More details about bearings can be found in section 3.1.2, and retrofitted corbels are dealt with in section 3.2.6. There are new developments regarding the optimisation of the reinforcement. In particular, the use of double-headed studs has alleviated the problem of anchoring the reinforcement properly. Double-headed studs (Fig. 2.158) enable the anchorage to be positioned exactly below the bearing. An overdesign so that the anchorage length can be reduced is no longer necessary, and that usually means less reinforcement is required. Experimental studies have verified the familiar failure through the reduction in the area of the compression zone at the base of the corbel. The load-carrying capacity is therefore equivalent to that of a corbel with conventional reinforcement. A National Technical Approval has already been issued for one product [148]. One practical advantage of double-headed studs is that it is possible to increase the size of the corbel at a later date. Such threaded studs also reduce the complexity of the moulds, and providing a shear key means we can assume a monolithic joint between corbel and column. The effective structural depth therefore begins at the lowest key within the depth of the column. Bracket solutions familiar in structural steelwork have appeared recently as corbels in precast concrete structures (Fig. 2.159). The cast-in items fitted flush with the face of the concrete are completed on site with a steel bracket and result in a completely concealed corbel. Fire protection is guaranteed by filling the joint with grout. Design is carried out according to the manufacturer’s specification. The design of notched beam ends is described in detail in [157, 158, 193–195]. According to those publications, the two truss action models shown in Fig. 2.160 can be used for the

150


Fig. 2.158 Corbel with double-headed studs (Halfen system)

Fig. 2.160

Fig. 2.159

Steel corbel (Peikko system)

Truss action models for notched beam ends

design. Care should be taken to ensure adequate anchorage of the reinforcement beyond the nodes of the truss model. A notched beam end functions in a similar way to a frame corner with a positive moment. In both cases diagonal reinforcement to limit crack propagation as a result of the high notch stresses in the uncracked condition is normally the most effective way of controlling the usually very early cracking in the internal corner of the notch. Often, the most appropriate approach is to employ a – from the constructional viewpoint – combined truss action model. In practice it is usual to choose truss action model (a) for moderate loads and a combination of the two for heavier loads. The authors recommend assigning 60 % of the load to each model and installing combined reinforcement. The minimum depth of the notch can be estimated by limiting the strut to min dk j

4 Ad b fcd

(35)

However, it is normally the anchorage length required and the space necessary for the reinforcement that govern the depth of the notch.


Fig. 2.161

151

Measured suspension force (taken from [157])

The design of notched beam ends is based on truss action models in [75], too. Concerning the question of whether TV w A is adequate for the design of the support reinforcement in truss action model (a), the reader is referred to the tests described in [157], in which the steel stresses in the vertical and diagonal suspension reinforcement were measured. The force to be suspended was smaller than the support reaction A (Fig. 2.161) in all the tests carried out. The reason for this was given as the potential supplementary arching effect corresponding to Fig. 2.162. Of course, important with truss action model (a) is that the reinforcement for TH in the bottom of the beam nib does not end in the beam before the first diagonal strut C3, but rather is anchored towards the centre of the beam from the point at which it intersects this strut. It is then – as the tests have shown – adequate to design TV for the support reaction A only without any surcharge. The additional transverse tensile forces due to anchoring the force TH are then taken by the shear links further inside. The normal shear force design is adequate for these links. With a diagonal suspension according to truss action model (b), TD w A/sin a.

152

Fig. 2.162 effect


Supplementary arch loadbearing

Fig. 2.163 Arrangement of reinforcement at the end of a heavily loaded beam

In truss action model (a) the horizontal tensile force TH is as follows: TH w

A L1 SH zk

(36)

where L1 distance from centre of bearing to centre of gravity of suspension reinforcement zk w 0.78 · dk The values for L1 and zk should be estimated carefully because the theoretical anchorage points depend on the reinforcement arrangement chosen and production and assembly tolerances must be taken into account as well. If a shear connector or any other type of fixing is chosen which can transfer the restraint forces, then the horizontal force TH, similarly to a corbel, should be increased with H w 0.20 · A. Where the suspension reinforcement is to be in the form of shear links only, then it is best to position these at an angle with respect to the notch. This arrangement has several advantages over vertical shear links: it reduces the nib force TH and gains more anchorage length at the bottommost beam truss node for anchoring T. Following on from the beam ends investigated in [157], two further beams corresponding to Fig. 2.164 were investigated which revealed good behaviour under service and ultimate loads. However, it is better to have straight ends to the bottom beam reinforcement and provide additional horizontal loops with appropriate lap lengths for the anchorage (similar to Fig. 2.163). The inclined shear links are somewhat longer than the vertical links and therefore must be given their own number on the reinforcement layout drawing. It is therefore also possible to use bars with a larger diameter in order to minimise the number of links required in the vicinity of the internal corner of the notch. Special solutions have been worked out for very low nib depths and for composite forms of construction in particular, where a cast-in steel part functions as the beam nib. Fig.

153


Fig. 2.164 Shear links in a notched beam end

Fig. 2.165 Special steel beam nib (Pfeifer system)

2.165 shows an example of how a support detail with such a steel nib functions. The suspension for the shear force is provided by the shear links and a double-headed stud which transfer the force to the cantilevering steel beam that transfers the load to the support. The design is carried out according to the manufacturer’s specification and approval documentation. As Fig. 2.165 indicates, the arrangement of the reinforcement must be worked out in detail. An angled notch is used only rarely these days. It is important here to choose the angle such that theoretically virtually no shear stresses occur, which means that shear reinforcement in the angled part of the web is no longer required. This type of support detail has advantages when it comes to routing building services (see section 2.5), but also the big advantage that the beam is loaded directly from above and the load does not need to be suspended from above via a continuous boot at the bottom. From the constructional viewpoint, however, it is important to ensure an extremely good anchorage for the tensile bending reinforcement because of the shallow-angled strut. In most cases it will be necessary to provide anchor plates to ensure a good anchorage. Continuous boots or individual corbels are very common in conjunction with inverted Tbeams and perimeter L-beams (Fig. 2.166). With such beam types the load on the boot must be suspended within the web of the beam. Where there is a boot on both sides and

154

Fig. 2.166


Loads applied to bottom of beam

a symmetrical loading (uniform load), suspension reinforcement for the total load applied at the bottom is frequently provided in practice in addition to the reinforcement required for shear. However, additional reinforcement for just 50 % of the suspended load is adequate, apart from in the vicinity of point of zero shear, provided the “offset” of the shear force resulting from a detailed consideration of the truss action model is also taken into account (see [75]). A one-sided boot supporting a uniformly distributed load represents a similar case. Here, assuming an additional force of a (37) DT w F 1S b is usually well on the safe side. In this case, too, the fact that part of the force has already been taken into account in the usual design of the section for shear and torsion is ignored. For simplicity, Fig. 2.166b ignores the fact that a part of the torsion moment is resisted by a horizontal couple resulting from the closed shear flow in the equivalent hollow crosssection. These influences are dealt with in [159] but assuming for simplicity that the system lines of the equivalent hollow cross-section are identical to the centroid axes of the shear links. Here, the additional force DT with respect to the design of the section for shear and torsion is specified as follows: 5 3a jF (38) DT w F S 8 4b This value is valid for the limiting case of a deep L-beam with z/h f 0 (Fig. 2.166c) and lies on the safe side for other z/h ratios. Individual corbels or point loads on a boot should be treated differently to this. In such situations the width over which the load is applied must be considered and the “suspension reinforcement” concentrated in the vicinity of the effective loading zone. The total suspended load can be taken as a DT w F 1S (39) b


155

Fig. 2.167 Perimeter beam with continuous boot (bearing pressure sk I 0.08 fck)

which lies on the safe side. However, in the zone in which this concentration of reinforcement is effective, the shear reinforcement resulting from the shear force and torsion moment components of the load under consideration do not have to be included additionally. With such boots it is important to realise that the normal reinforcement arrangement contradicts the principles given above for notched beam ends in some respects (Fig. 2.167). For example, it is not usual to have a horizontal loop below the bearing pad, but rather a vertical link, and the bearing pad or bearing strip is positioned at a distance of only c2 from the edge of the boot. This is possible with maximum bearing pressures up to about sk I 0.08 fck. But the resultant of the support reaction (taking into account potential tolerances) must be applied within the top longitudinal bar in the boot such that a certain clearance of about c1 remains before the start of the bend in the link. Further, the direction of the bend in the web link does not correspond to the requirements of the suspension reinforcement for notched beam ends. This means that the inclined compressive force C in the boot is supported on the bottom longitudinal reinforcement in the web and a correspondingly small inner lever arm z must be chosen. In the case of heavily loaded separate corbels at the bottom edges of beams, providing horizontal loops in the corbels below the bearing pad according to Fig. 2.157 and additional loops to suspend the load in the web are unavoidable. A floor slab, e.g. in the form of double-T units, must be connected rigidly to a perimeter beam as shown in Fig. 2.168 if torsion in the latter is to be avoided in the final condition. The upper compression contact is achieved by a grout filling, the lower tensile force transferred via a dowel, surrounded by loops, or – in the case of heavier loads – by welding together anchor plates cast into the web of the double-T unit and the boot, or via screw couplers. The dowel is cast into the boot and fits into a corrugated sleeve in the web of the double-T unit, which is filled with grout after erection. When using inverted channel section floor units, the tension reinforcement can be laid in the grouted joint and anchored to the perimeter beam via a screw coupler. When planning boots on the bottoms of beams, it is essential to consider the fact that the beam must be supported temporarily during construction unless it has been designed for torsion loads.

156


Fig. 2.168

Rigid connection between floor slab and perimeter beam

2.6.3

Lateral buckling

The slender beams or rafters to the roofs of single-storey sheds so common in precast concrete construction must be checked for lateral buckling, i.e. the lateral stability of the compression flange during demoulding, storage, transport and erection as well as the final condition. Ref. [161] contains a detailed overview and assessment of practical methods for assessing lateral buckling. Based on that publication, Table 2.9 contains the basic equations for the lateral buckling moment for an ideal elastic material for the rectangular, T- or I-sections with one or two axes of symmetry and Iy II Ix so customary in reinforced and prestressed concrete construction. The position of the application of the load is approximated to the shear centre for this table. The actual relationships with the load application point of the dead load at the centre of gravity or an additional load applied to the top flange result in values with a scatter of e10 % for slender roof beams. A figure of 0.4 is assumed for the following relationship: G=E w

1 2 (1 S m)

i.e. a Poisson’s ratio of m w 0.25, which is appropriate for the high concrete qualities normally used in precast concrete construction. The critical lateral buckling loads for duopitch roof beams should be reduced compared with those for beams with a constant depth. The reduction factors given in Table 2.10 can be used within the scope of the simplifications assumed here [163].

157


Table 2.9 Lateral buckling moment for ideal elastic material, forked support and T- or I-cross sections with single or double symmetry and Iy II Ix Mk Loading

k1 pffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi k1 E pffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi 0,4 Iy IT EIy GIT l 1 max M

k1

M

p

1.00

q l2 8

3.54

1.12

Pl 4

4.23

1.35

jw

k1 p

Table 2.10 Reduction factors h for determining the lateral buckling load for duopitch roof beams [163]

Cross-sectional form

Reduction factor h for dA/dM ratio 1

0.75

0.5

0.25

Rectangular, any d/b value

hw1

0.87

0.74

0.61

I-section symmetrical

hw1

0.96

0.82

0.73

about two axes

Crane slings for lifting roof beams usually have to be attached to two intermediate points (see [3]). The risk of lateral buckling therefore decreases as the suspension height increases. The best solution is to support the beam roughly at its quarter-points because lateral buckling is then impossible. However, it must be remembered that the elasticity of the crane sling means that only a reduced suspension height can be effective.

158


In the case of reinforced and prestressed concrete beams – the stress-strain graph is not linear, – the flexural and torsional stiffness depends on the loads, especially at the transition to the cracked condition, and – the beam is cast with certain imperfections, which means that the aforementioned equations should only be used in conjunction with appropriate safety factors, generally with g w 4.0 –5.0. According to Stiglat [166], the lateral buckling moment MK of the beam made from an ideal elastic material is therefore reduced, according to Table 2.9, to MlK w

sT MK sT W0 sK

(40)

where sK w

MK W0

W0 moment of resistance at the upper compressed edge of cross-section sK stress in extreme fibre at upper compressed edge of cross-section due to MK in uncracked condition sT stress in a theoretical bar in buckling with the same slenderness lv as the buckling beam The comparative slenderness lv is calculated as follows: rffiffiffiffiffi Eb lv w p sk where Eb is the characteristic value of the elastic modulus of the concrete to DIN 1045:1988. In reality, however, the elastic modulus depends on the loading, as given by the curved stress-strain diagram for the concrete. The stress curves for sT (Figs 2.169 and 2.170) are therefore drawn according to the realistic data given in [167], based on the tangent modulus in this case. Using the large-scale tests on reinforced and prestressed concrete beams carried out in the meantime [179], Stiglat has confirmed the adequacy of the accuracy of his simple method and therefore regards a global factor of safety of g w 2.0 as adequate [181]. The large-scale tests of Ko¨nig and Pauli [179] resulted in a method of calculation they describe in [180]. As this serves as a starting point for the majority of computer design programs, the main concepts will be explained below. The underlying idea is to verify a potential state of equilibrium in the deformed system. In doing so, a figure of twice the initial deformation 40 is used in the calculation for simplicity instead of the creep deformation.


159

Fig. 2.169 Analysis of lateral buckling according to Stiglat [166]; approach using the fictitious elastic modules Ef for various cross-sectional forms

Fig. 2.170 Analysis of lateral buckling according to Stiglat [166]; tT and lV values

1. Limit state consideration: the potential twisting of the beam is limited by the moment that can be accommodated about the weak axis of the cross-section. 4Biegung w 4ges. s 40 Mz,Rd 4Biegung w s 40 My,Sd

(41)

2. Limit state consideration: the potential twisting of the beam is limited by the maximum torsion moment that can be accommodated without reinforcement, which corresponds to the cracking moment. ðl 4Torsion w

MT (x) MT (x) dx GIT (x)

0

where max MT

w MT,Riss w fctm WT

(42)

160

Fig. 2.171


Equilibrium in the deformed system

Fig. 2.172

Deformed position of beam

The potential limit twist of the beam therefore represents a cross-sectional value made up of the following components: (x) 4 4grenz (x) w 40 S min Biegung (43) 4Torsion (x) These are compared with the real deformations due to external loads. If these are now smaller than the limit deformation that can be accommodated by the cross-section, then stability against lateral buckling can be regarded as guaranteed. In their publication [179], Ko¨nig and Pauli scrutinise the formula given in EC 2-1-1 section 4.3.5.7 in which stability against lateral buckling is considered to be adequate when l0 J 50 b h J 2,5 b

(44)

where l0 distance between lateral supports b width of compression flange h depth of beam Their findings show that a beam can be regarded as at risk of lateral buckling as soon as the load-carrying capacity for biaxial bending using second-order theory is reduced by

161


Fig. 2.173 Series of calculations for beams at risk of lateral buckling [180]

Fig. 2.174 Non-rigid forked support for resisting lateral buckling

Fig. 2.175 This top flange was obviously a bit too narrow!

more than 10 % compared to the acceptable ultimate moment as a result of primary bending. The following empirical formula has been derived from this: ffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi s l0 3 4 h (45) bj 50 (see Fig. 2.173; also DIN 1045-1 section 8.6.8 (2)). As a result of this work, the DAfStb application guidelines for EC 2 part 1 have reduced the above condition to l0 J 35 b h J 2,5 b

(46)

Mann (in [168] and [169]) attempts to attribute the lateral buckling problem of a slender reinforced concrete roof beam to the buckling of the top flange due to compressive bend-

162


ing forces. The actual verification of lateral stability is confined here to an additional analysis of the safety of the beam at the ultimate limit state. This beam has a top flange width b w vb (v w reduction factor) which depends on an ideal slenderness l, an ideal eccentricity e and the reinforcement in the compression flange m0, and can be designed using the design tables for slender compression members or with the help of appropriate computer programs, for example. The reader is referred to [161] and its extensive bibliography for details of the lateral buckling analyses for reinforced concrete beams according to Rafla or Ro¨der/Mehlhorn, which are very complicated to use in practice. Mehlhorn, Ro¨der and Schulz describe an approximation method with the help of an ultimate limit state analysis for biaxial bending in [176] where they use examples that use the partial safety factors according to the Eurocode. This method is based on [177], which is compared with another solution in [178]. An attempt to derive a sufficiently accurate rough verification of stability against lateral buckling, based on Rafla, can be found in [170]. Mattheiß specifies a method of estimating a laterally stable compression flange width in [186]. Forked supports for beams can be constructed according to the typical details for frame construction shown in Figs 2.176a and 2.177a. With the types of support shown in Figs 2.176b and 2.177b, the lack of shear resistance in the elastomeric bearings makes it neces-

Fig. 2.176 Support details for T-section beams (Fachvereinigung Deutscher Betonfertigteilbau e.V.)

Fig. 2.177 Support details for I-section beams (Fachvereinigung Deutscher Betonfertigteilbau e.V.)


163

sary to include shear connectors to resist horizontal forces. The horizontal forces therefore tend to concentrate around the upper cut-out, which must be taken into account when designing the connecting part. With a form of lateral restraint according to Fig. 2.174, it may be necessary to take into account the spring stiffness of the fork, as shown in [171]. According to DIN 1045-1 section 8.6.8, the fork should be designed in such a way that it can handle the following torsion moment: Td w Vsd

leff 300

(47)

where Vsd vertical design shear force at support leff effective support width of beam How the stiffness of the fork affects the lateral buckling behaviour has been investigated in [187]. The shear links at the end of the beam and the end anchorage of the longitudinal reinforcement must be arranged in such a way that this torsion moment can be accommodated, i.e. the link closer bars must be designed with laps. Furthermore, it should be remembered that the torsion moments must be analysed right down to the foundations. 2.6.4

Pad foundations

Whereas pocket foundations are described in [3] and [189] (see also [190]), only the pad foundations that have become more popular in recent years because of their economy will be dealt with here (Fig. 2.178). An appropriate shear key on the base of the column and the sides of the pocket enables pad foundations to function as if the foundation had been cast monolithically with the column. This has been confirmed by tests (see [172], also [99]). The embedment depth of the column in the foundation should be at least t w 1.5 · c. The thickness of the foundation below the column then depends on the structural depth of the foundation d required or the punching analysis for the base of the foundation for the temporary condition during construction, i.e. pocket not yet filled with grout. The width of the pocket should be equal to c S 2df. In order to accommodate the tolerances and to enable the precast concrete column and the grout to be installed properly, the grouting space around the column should be approx. 7.5 cm wide, which results in a pocket width of bpocket w c S 15 cm. The shear key is formed either by permanent corrugated sheet metal formwork, with a corrugation depth of i 1 cm, or with timber battens (Fig. 2.95) matching the requirements shown in Fig. 2.179 (see also Fig. 3.34). The design value for bond stress may be increased by 50 % when calculating the anchorage lengths of column reinforcement with straight ends embedded in a foundation (see DIN 1045-1 section 12.5(5)). In doing so, there must be concrete cover guaranteed on all sides by shear links and a transverse pressure at 90h to the plane of the reinforcement. According to DIN1045-1 section 12.4(2), good bond conditions may be assumed for horizontally cast column cross-sections with c J 50 cm.

164

Fig. 2.178 Pad foundation showing arrangement of reinforcement


Fig. 2.179 Distribution of moments and arrangement of reinforcement for axial loads

The standard design of the foundation can be carried out separately for the load components due to axial force and bending moment. The design for the axial force component can be carried out for the section along the edge of the column (Fig. 2.179): NSt b c 2 MBem 1s (48) w N 8 b which has been confirmed by tests and is also valid for monolithic foundations. The bending reinforcement for the axial force component may be distributed uniformly over the width in the case of a small foundation width. But if b i c S d, then the bending reinforcement should be graduated to match the bending moment diagram. The design for bending caused by the moment component MSt is resisted by an “equivalent beam” of width b1 w c S d (Fig. 2.180). The resulting amount of reinforcement AM s is distributed over a width of b2 w 0.5 · b1 and bent up behind the pocket in the foundation to function as vertical starter bars. The horizontal reinforcement required, which depends on the offset of the bent-up foundation reinforcement and the tension reinforcement in the column, is identical for reasons of equilibrium:

165


Fig. 2.180 Equivalent beam for accommodating the column moment

a M AH S w AS l

(49)

where a distance between vertical starter bars and longitudinal reinforcement in column l difference between embedment depth and anchorage length (Fig. 2.181) This is a type of frame corner with two frame legs with different cross-sectional dimensions where the peripheral ties must be suitably connected. According to [172], the punching shear analysis in the fully grouted state can be carried out in the same way as for a monolithic foundation according to DIN 1045-1. An angle of 34h should be assumed for the punching cone; according to DAfStb booklet 525, this angle should be increased to 45h in the case of stocky foundations. In that case 100 % of the bearing pressure acting in this circular cross-section can be deducted from the punching load. Likewise, the shear capacity vRd,ct may be increased in relation to the circular cross-sections ucrit, 1.5d/ucrit, 1.0d.

Fig. 2.181 pocket

Truss action model in vicinity of

Fig. 2.182 Effective ultimate shear stresses plotted against slab slenderness [172]

166


Fig. 2.183 loading

Punching shear analysis for eccentric

However, the tests have revealed that the necessary safety against punching is not quite attained in stocky foundations with 1 2d J 1,0 (50) 0,75 J w l (bsc) The reason for this is the lower transverse pressure acting on the shear joint in the top third of the joint. Initially, vertical shear planes running from top to bottom appeared in very stocky foundations, which only manifested themselves as a diagonal, outward shear crack at a certain depth. Designers are therefore recommended to reduce the shear capacity vRd,ct by a factor of 2.2–1.7 · (1/l) in the punching shear analysis for the above slenderness range. A separate analysis of punching shear for the self-weight of the column, acting via its base plate, must be carried out for the base of the pocket for the temporary condition during construction. A simple method for checking punching shear for an eccentric load is given in [172] (Fig. 2.183). According to this, the shear stresses are determined for the most heavily loaded quarter of the plate. The resulting shear force here is the content of the stress body cut off at an angle reduced by the bearing pressures on the associated quarter of the area of the punching cone. According to EC 2, increasing the critical shear force by a factor b w 1.4, as with perimeter columns, and then designing the foundation as though it were loaded concentrically is an acceptable approximation for taking into account an eccentric loading. The reader is also referred to [173] and [192] for the analysis of punching shear.


2.6.5

167

Design for fire

The sizing of components for fire protection purposes can be carried out with the help of a fire loading case calculation (thermal analysis) or by using simplified comparative data according to DIN 4102- 4. The former can be carried out with the help of DIN EN 1992-1-2:2006. A number of computer programs for automating this thermal analysis are currently undergoing development. Ref. [198] contains an up-to-date overview of the valid regulations and minimum dimensions according to DIN 4102- 4. The verification by way of comparative dimensions according to DIN 4102- 4 was the standard method used in practice in recent years. However, the inclusion of DIN 10451 in building legislation made it necessary to adapt DIN 4102- 4, mainly because of the altered safety concept and level of design, which in some cases calls for higher degrees of utilisation for the materials and the consideration of high-strength concrete up to grade C80/95 [199]. DIN 4102-22 was therefore drawn up to provide a so-called application standard. At the moment the sizing of components for fire events is undergoing changes because of the changeover in both German and European standards, a fact that is discernible in the number of regulations and publications. Further substantial changes will take place here in the coming years. A number of the provisions of DIN 4102- 4 important to the design of precast concrete components for fire, taking into account amendment DIN 4102- 4/A1 and DIN 410222, are summarised below. The behaviour of components made from reinforced or prestressed concrete when exposed to fire, and hence the fire resistance rating, essentially depends on the following factors: 1. Dimensions of component (cross-section, slenderness, distances between bar axes and edges) 2. Type of exposure to fire (one or more sides) 3. Building materials (type of steel, concrete aggregates) 4. Structural system (statically determinate or indeterminate support, loads carried via one or two axes) 5. Construction details of supports, connections and joints 6. Degree of utilisation of concrete and steel strengths 7. Additional protective measures (render, plaster, cladding, suspended ceiling, lining) In typical multi-storey buildings, i.e. more than two storeys but not high-rise buildings, it is generally sufficient when the building materials comply with class B2 (flammable) as a minimum and the joints separating walls between buildings and layers of insulation in the façade comply with class B1 (not readily flammable). On the other hand, walls, columns, floors and stairs with loadbearing and/or stability functions must generally have a fire resistance rating of F 90 -A. Rating F 30 is generally adequate for buildings with less than two storeys, whereas F 120 is necessary for high-rise buildings, and even F 180 for buildings more than 200 m high.

168


Materials in expansion joints must comply with class A (incombustible). According to [45], there are no objections to using materials of class B2 for elastomeric bearings in buildings in which statically determinate support conditions generally apply to the precast concrete components. Non-combustible materials or fire-retardant forms of construction (F 30 -B) are required for non-loadbearing room-enclosing external walls (also spandrel and fascia panels). The most common building authority requirements are therefore F 30 -A and F 90 -A. Special attention should be given to the design and construction of fire walls or multilayer separating walls. In the case of the latter, it is a requirement that the wall should remain stable when exposed to a fire from the left or the right side and simultaneously withstand horizontal loads due to wind, bracing forces and impact [200]. In the normal case reinforced concrete components designed to DIN 1045 meet the F 30 A requirement. Certain minimum cross-sectional dimensions and distances u between centre of reinforcement and edge of concrete are necessary in order to achieve the F 90 -A rating. The minimum edge distances given below therefore always refer to the distance between the centre of the reinforcing bar concerned and the surface of the component, and not the concrete cover, which is measured from the surface of the reinforcing bar to the face of the concrete. The minimum dimensions of the individual elements for F 30 and F 90 ratings for typical elements in frame structures are given in section 2.3. The fire resistance of reinforced concrete components can be increased if necessary by applying coats of suitable renders and plasters that exhibit a good bond with the concrete. This option can be particularly attractive for floor slabs. The beams (Fig. 2.184) encountered in precast concrete construction generally have statically determinate support conditions. Table 2.11 summarises the minimum widths and minimum reinforcement edge distances for reinforced and prestressed concrete beams for various fire resistance ratings and the usual three-sided exposure to fire. As prestressing steel is generally more sensitive to fire loads, the strands should be positioned more towards the centre of the component, whereas the reinforcing steel in conventionally re-

Fig. 2.184

Beams

169


Table 2.11 Minimum widths and minimum reinforcement edge distances for reinforced and prestressed concrete beams to DIN 4102-4 Fire resistance rating F 30-A

F 60-A

F 90-A

F 120-A

80

120

150

200

Min. width b in mm of non-clad prestressed concrete 120 beam 1) in tensile bending zone or precompressed tension zone 2)

160

190

240

Min. web thickness t in mm of non-clad beam in tensile 80 bending zone or precompressed tension zone 2)

90

100

120

Min. edge distances u and us in mm of tension reinforcement in non-clad reinforced concrete beam with one layer of reinforcement for a given beam width b in mm

b w 80 u w 25 us w 35

b w 120 u w 40 us w 50

b w 150 u w 55 4) us w 65

b w 200 u w 65 4) us w 75

b w 160 u w 10 us w 20

b w 200 u w 30 us w 40

b w 250 u w 40 us w 50

b w 300 u w 50 4) us w 60

b w 120 u w 30 us w 40

b w 160 u w 50 us w 60

b w 200 u w 60 4) us w 70

b w 240 u w 70 4) us w 80

b w 160 u w 25 us w 35

b w 200 u w 45 us w 55

b w 250 u w 55 4) us w 65

b w 300 u w 65 4) us w 75

Min. width b in mm of non-clad reinforced concrete beam in tensile bending zone

Min. edge distances u and us in mm of tension reinforcement in non-clad prestressed concrete beam 1) with one layer of reinforcement for a given beam width b in mm 3)

1)

Prestressing wires or strands according to National Technical Approval. DIN 4102-4 Table 4 must be taken into account in the compression or compressive bending zone or in the precompressed tension zone at the supports. 3) The Du-values for strands and wires to DIN 4102-4 Table 1 have been taken into account (Du w 15 mm). 4) With a concrete cover c i 50 mm, additional reinforcement according to DIN 4102-4 section 3.1.5.2 is required in the cover. 2)

inforced elements is arranged around the perimeter of the component. Additional reinforcement within the cover is necessary when the concrete cover c exceeds 50 mm. Standard, non-continuous plastic bar spacers do not have an influence on the fire resistance rating [201]. On the other hand, cast-in channels can have an effect on the reaction to fire of a reinforced concrete component. The u-values required must be verified by a test certificate. The minimum cross-sectional areas and edge distances for corbels and notched beam ends are given in Fig. 2.185. Joints between components with a width a J 30 mm can be ne-

170


Fig. 2.185 Minimum cross-sectional areas for corbels, notched beam ends and beam openings

glected and the component surfaces within the joint can be regarded as being not exposed to the fire. At openings in webs, the remaining cross-section of the tension flange should be i 2 b2min. Openings with a diameter I 100 mm may be ignored. The minimum depth of reinforced or prestressed concrete floor slabs with no plaster to the soffit should be d j 100 mm in order to comply with the F 90 requirements. This value also applies to the total thickness D of floor slabs with an incombustible bonded screed, although in that case the depth of the precast concrete floor unit must be d j 50 mm and that of the screed dE j 25 mm. Table 2.12 shows the minimum slab dimensions and the minimum reinforcement edge distances. In hollow-core slabs, Anet/b i 100 mm and the minimum distance between bottom of void and soffit must be du j 50 mm. The minimum edge distance of the span reinforcement for simply supported solid or hollow-core slabs is u w 35 mm for F 90 (see also Table 2.12).

Fig. 2.186

Floor slabs to DIN 4102-4

171


Table 2.12 Minimum depths and minimum reinforcement edge distances for solid reinforced and prestressed concrete floor slabs to DIN 4102-4 Fire resistance rating F 30-A

F 60-A

F 90-A

F 120-A

Min. depth h in mm of non-clad solid slab without screed for statically determinate and indeterminate support conditions

60 2)3)4)

80 2)

100

120

Min. edge distance u of span reinforcement in reinforced concrete slabs 1) without load-carrying effect in transverse direction

10

25

35

45

10 10

10 25

20 35

30 45

Min. edge distance uo of support or fixity reinforcement 10 in reinforced concrete slabs 1) without load-carrying effect in transverse direction

10

15

30

Min. edge distance u of span reinforcement in reinforced concrete slabs 1) with load-carrying effect in transverse direction and the following ratio: b/l J 1.0 b/l J 3.0

1)

The u-values must be increased by the Du-values according to DIN 4102-4 Fig. 1 in the case of solid prestressed concrete slabs. 2) The minimum depth of slabs exposed to fire on more than one side (e.g. cantilevering slabs) must be h j 100 mm. 3) The minimum depth for statically indeterminate supports must be h w 80 mm. 4) According to DIN 1045-1 section 13.3, the minimum depth of solid slabs is h w 70 mm.

DIN 4102- 4 does not mention prestressed hollow-core slabs. But applying the provisions correspondingly would result in a requirement of u w 50 mm when using prestressing strands of grade St 1570/1770, which would require additional reinforcement within the depth of the cover. A lower u-value can only be achieved by using carbonaceous aggregates or additional reinforcement (see [45]), or by reducing the permissible stresses in the prestressing steel. This is one of the reasons why this type of floor slab is not in widespread use in Germany.

Fig. 2.187 Reinforcement in precast concrete floor planks with a structurally effective in situ concrete topping (example according to [45])

172


Table 2.13 Minimum sizes and minimum reinforcement edge distances for reinforced concrete columns to DIN 4102-4 Fire resistance rating F 30-A

Min lcol w 2,0 m Max lcol w 6,0 m

F 60-A

F 90-A

F 120-A

Min lcol w 1,70 m Max lcol w 5,0 m

Min. cross-sectional dimensions of non-clad reinforced concrete columns 1) for exposure to fire on more than one side and a utilisation factor a1 as follows: Utilisation factor a1 w 0.2 Column length min. lcol min. size h in mm associated min. edge distance u in mm

120 34

120 34

150 34

180 37

Column length max. lcol min. size h in mm associated min. edge distance u in mm

120 34

120 34

180 37

240 34

Column length min. lcol min. size h in mm associated min. edge distance u in mm

120 34

160 34

200 34

260 46


120 34

180 37

270 34

300 40

Column length min. lcol min. size h in mm associated min. edge distance u in mm

120 34

190 34

250 37

320 40


120 34

250 37

320 40

360 46

Utilisation factor a1 w 0.5

Utilisation factor a1 w 0.7

1)

Minimum dimensions for compression members within helical or horizontal bars provided no higher values are specified: F 30: h w 240 mm, F 60 to F 120: h w 300 mm

In composite plank floors the planks must also be at least 50 mm deep in order to meet the F 90 requirements. An average um value can be calculated from the positions of the additional reinforcing bars, the mesh reinforcement in the plank and the longitudinal bars of the lattice beam, which must then be i 35 mm (Fig. 2.187). In the case of reinforced concrete columns (Table 2.13) of minimum size it is the length and the degree of utilisation that are critical. The utilisation factor is the ratio of the axial force

173


Table 2.14 Minimum thicknesses and minimum reinforcement edge distances for reinforced concrete walls to DIN 4102-4 Fire resistance rating F 30-A

F 60-A

F 90-A

F 120-A

80

90

100

120

80 100 120

90 110 130

100 120 140

120 150 160

10

10

10

10

10 10 10

10 10 10

10 20 25

10 25 35

Non-clad walls 1) with permissible slenderness (w storey height /wall thickness w hs/h) to DIN 1045-1 Min. wall thickness h in mm for: Non-loadbearing walls Loadbearing walls with: utilisation factor a1 w 0.07 utilisation factor a1 w 0.35 utilisation factor a1 w 0.70 Min. spacing u of longitudinal reinforcement in mm for: Non-loadbearing walls Loadbearing walls with: utilisation factor a1 w 0.07 utilisation factor a1 w 0.35 utilisation factor a1 w 0.70 1)

Reductions are possible for walls with render/plaster on both sides according to DIN 4102-4 sections 3.1.6.1 to 3.1.6.5; however, the minimum wall thickness is h w 60 mm for non-loadbearing walls, h w 80 mm for loadbearing walls.

present in the fire loading case to the load-carrying capacity of the design condition (Nfi,d/ NRd). As a result of the reduction in the loads and the factor of safety in the fire loading case, a utilisation factor of 0.7, for example, means that the column is generally 100 % utilised in the design condition. However, Table 2.13 applies only to columns restrained against rotation at both ends in braced buildings. Furthermore, Table 2.13 can only be used for the column lengths given in the table. One remedy for columns pinned at one end is to carry out the design with a longer buckling length. Up until now it was not possible to use simplified methods of calculation for columns in the form of a vertical cantilever. But a simplified method of analysis with which such columns can be designed for the fire loading case has been developed in [351]. The method covers the normal range of applications and can also be used for laterally restrained columns by selecting a suitable buckling length. In the case of fully utilised room-enclosing walls exposed to fire on one side and with a slenderness according to DIN 1045-1, the minimum thickness of a wall complying with F 90 requirements is h w 140 mm and the minimum reinforcement edge distance u w 25 mm (Table 2.14). Lower values apply if the wall is not fully utilised. The same values apply when a fire wall is loadbearing, except that the slenderness is limited to hs/d I 25. For a loadbearing multi-layer separating wall the minimum thickness of the wall should be h w 300 mm and the minimum reinforcement edge distance u w 55 mm (see also [198]).

174

Fig. 2.188


Joints between precast concrete elements

Segmented walls with doors and windows are dealt with in DIN 4102- 4. The reader should also refer to [188] for information on the fire protection analysis of pier cross-sections in fenstrate façades.

Joints [175] between precast concrete floor units must be filled with mortar or concrete in accordance with Fig. 2.188. From the fire protection viewpoint, joints up to a width of 3 cm may also remain open if the floor units are provided with an in situ concrete topping according to Fig. 2.186c. Joints between ribs must be closed off with mortar as shown in Fig. 2.188b. The relevant width b necessary for determining u and us may be related to both ribs in the case of a joint width I 2.0 cm. Joints between precast units in roofs may also remain open up to a width of 2 cm provided a layer of thermal insulation i 8 cm thick complying with building materials class A is laid on the top of the units.

3

Joints between precast concrete elements

Whereas one of the key features of in situ concrete is that the structure is “built from one mould” as it were, in precast concrete construction the individual, prefabricated parts are only assembled to form a structure at a later point in time. The joints between the individual elements must therefore transfer the forces in some appropriate way. A summary of joints in precast concrete construction can be found in [202] and [74]. Although in many instances joints have to be designed to accommodate axial forces, shear forces and bending moments, compression, tension and shear joints will be treated separately in this chapter. 3.1

Compression joints

3.1.1

Butt joints

Precast concrete components should always be bedded on bearings or a layer of mortar [203, 204, 224]. Dry bearings without intermediate layers should not be used. According to DIN 1045-1 section 13.18.2, dry bearings are only permissible when “the average compressive stress in the concrete does not exceed 0.4fcd and the necessary quality of workmanship is achieved in the factory and on the building site” (e.g. intermediate components in floors or roofs). However, according to current practice in Germany, a pad of insulating board or similar material should always be provided as a very minimum in all such bearings. DIN 1045-1 makes a distinction between joints with soft and hard bearings. In the case of a joint with a soft bearing, (Fig. 3.1a), the lateral displacement of the jointing material leads to tension forces in the end faces. The resulting lateral tensile stresses must be resisted by reinforcement. A joint with a soft bearing may even require reinforcement in the joint itself. A joint with a hard bearing is a type of joint in which the elastic modulus of the jointing material is equal to at least 70 % of the elastic modulus of the adjoining component. A joint with a hard bearing plus reduced cross-section (Fig. 3.1b) gives rise to lateral tensile forces as a result of the redirection of the forces from the whole cross-section to the reduced cross-section, which has to be resisted by reinforcement [75]. Higher local bearing pressures are permissible in such situations (Fig. 3.2). According to DIN 1045-1 eq. (116), the following applies: FRdu w Ac0 fcd

pffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi Ac1 =Ac0 J 3,0fcd Ac0

(51)

Saleh has carried out further investigations [210]; a summary can be found in [209].


176

3 Joints between precast concrete elements

Fig. 3.1 Different types of compression joint: (a) soft bearing, lateral tensile stresses due to lateral spreading of jointing material, (b) hard constricted bearing, lateral tensile stresses due to reduction in transfer area, (c) hard non-constricted bearing, lateral tensile stresses due to redirection of load component in longitudinal bars and concrete encasement.

Fig. 3.2 Reduced area for determining local bearing pressure

A column joint with a hard bearing over the full area is essentially the normal case for heavily loaded column joints. The load-carrying capacity can be calculated using the following equation taken from DIN 1045-1: NRd w k Ac,n fcd S As fyd where 1,0 for steel end plate kw 0,9 for reinforcement in end face Lateral tensile stresses build up in the ends of the column elements directly adjacent to the joint as a result of the redirection of the load component in the column reinforcement and the concrete (see Fig. 3.1c). The investigations of Ko¨nig and Minnert [211] led to the publication of DAfStb booklet 499 with new design proposals for butt-jointed precast concrete columns made from high-strength concretes. For butt joints in normal-strength concrete, see also [212]. We basically distinguish between two types of butt joint detail (Fig. 3.3): – those with a steel plate, and – those with reinforcement in the end faces. The investigations have shown that using steel plates in the column end faces is a very efficient way of preventing lateral strains in the mortar joint and the resulting stresses tend to be low. Further, the total load component in the longitudinal reinforcement can be carried via the mortar joint so that there are no stresses due to the end anchorage of the longitudinal reinforcement (Fig. 3.3) in the vicinity of the joint.

3.1 Compression joints

177

Fig. 3.3 Compression of longitudinal reinforcement in a test column with and without steel plate on end of column [211]

When using reinforcement in the end faces, on the other hand, only a part of the force in the reinforcing bars is carried via end bearing. The larger part is transferred into the surrounding concrete via bond stresses. The higher concrete stress must be resisted by peripheral reinforcement at the base of the column above the joint (Fig. 3.4); adequate peripheral reinforcement is essential here. The lateral tensile stresses in the mortar joint must be resisted by the reinforcement in the column end faces. In working out the joint detail, care should be to taken to ensure that the reinforcement in the end face is built into the column directly without concrete cover and that the diameter of the bars does not exceed ds w 12 mm. The outer nodes of the mesh must be positioned on the outside faces of the column and the intersections must be carefully welded. The spacing of the bars must be J 5 cm and the shear links should be positioned as shown in Fig. 3.5.

178


Fig. 3.4 Effective constricted area in the plane of the shear link [212]

The following should used forllink : be fyd gs ds llink ¼ 0,75 4 flink;d where flink;d ¼ 2,25 fctk,0:05 In practice it seems that custom meshes for column end faces are not generally readily available. In addition, a maximum permissible joint thickness of 2 cm should not be exceeded for column butt joints with a hard bearing. Adhering to this maximum joint thickness is often difficult in practice because the manufacturing tolerances for suspended floors or ground slabs usually lead to thicker joints. The investigations of Paschen and Zillich [206, 207] led to the publication of DAfStb booklet 316 in which thicker joints are permitted as well because it was based on the old version of DIN 1045 (1988). The booklet also distinguishes between reinforced and unreinforced joints. The reduction factor k for determining the load-carrying capacity can be calculated according to Fig. 3.6 depending on the

Fig. 3.5

Details of column compression joint [212]


Fig. 3.6 Reduction factor k for the admissible design force of concentrically loaded column joints plotted against amount of longitudinal reinforcement and thickness of mortar joint (intermediate values may be obtained by linear interpolation)

179

Fig. 3.7 Additional transverse reinforcement at wall joints

joint thickness [206]. In doing so, however, the reduction factor should not exceed 0.9, which is in line with DIN 1045-1. The reduction factor k here is a function of the geometrical degree of reinforcement r and the joint thickness hj. The lateral tensile reinforcement necessary for this should be determined according to the acknowledged design methods. According to DIN 1045-1, in a joint where axial and shear forces act simultaneously, the shear forces can be neglected when VEd I 0.1 · NEd. When designing compression joint zones within wall joints supporting suspended floors on one or both sides, it is necessary to take into account lateral tensile stresses caused by rotation of the floor at the support. That can be dealt with in a simple way according to DIN 1045-1 section 13.7.2 by assuming that only 50 % of the loadbearing wall cross-section is used in the stress analysis for the wall above and below the joint. However, DIN 1045-1 states that 60 % of the loadbearing wall cross-section may be taken into account in the design if transverse reinforcement is provided in the wall above and below the joint (Fig. 3.7). This must be able to accommodate the following design tensile force at least: asw w h=8 asw in cm2/m, h in cm The spacing of the transverse reinforcement sw in the direction of the longitudinal wall axis must be h sJ 200 mm (the smaller figure governs)

180


and the diameter ds of the longitudinal reinforcement Asl at the base of the wall must at least 6 mm. A cross-section proportion i 60 % can then be considered if this is verified by test results that reproduce the true support conditions accurately [226]. 3.1.2

Zones of support to DIN 1045-1

DIN 1045-1 section 13.8.4 “Zones of support” does not deal with the bearings themselves, but rather the construction details of the bearing areas for suspended floors and beams. DAfStb booklet 525 contains more detailed information (available in German only). Besides the detailed design of the support zones, the following factors are also crucial when designing a support: – The dimensions of the reinforcement in the adjoining components – The maximum permissible bearing pressures – The choice of a suitable bearing EC 2 makes a distinction between “isolated members” and “non-isolated members”. The latter are components, e.g. hollow-core or solid slabs, that in the event of failure of the support can draw on loadbearing reserves from the transverse distribution of the loads, which is possible, for example, by grouting the longitudinal joints. Isolated members, e.g. roof beams or downstand beams, on the other hand, do not benefit from such a property. The length of the support (Fig. 3.8) is made up of the actual length of the bearing a1 and the allowances a2 and a3 which prevent spalling of the concrete in the supporting and supported components. In this situation the allowances on either side of the bearing are not added together, but instead linked statistically. See DAfStb booklet 525 for further information.

Fig. 3.8 Support zone: (a) elevation, (b) plan

Fig. 3.9 Horizontal support for a beam outside the plane of the bearing


181

The support length a may need to be longer where sliding bearings are in use. Likewise, in the case where a beam is not restrained horizontally in the plane of the support (Fig. 3.9), the gap t1 will need to be widened to allow for the effects of rotation about the joint. 3.1.3

Elastomeric bearings to DIN 4141

Ref. [213] contains an article on the introduction of the new standard on structural bearings, DIN EN 1337. As parts of this standard have yet to be published and many of the main relationships described below are of a technical nature and therefore not dependent on a standard, the design provisions according to DIN 4141 will continue to be used here. DIN 4141-3 “Structural bearings” divides supports into two categories. If the adjoining components, apart from the respective theoretical bearing pressure in the support joint, are not loaded to a significant extent by other support reactions and the stability of the structure is not at risk if the support is overloaded or the support function fails, then the support complies with the requirements of category 2. Category 1, on the other hand, covers all support conditions that must be verified by analysis, where failure or overloading of the support may involve a risk to the stability of the structure. Category 2 applies to the majority of instances in everyday precast concrete buildings for the supports to suspended floor slabs and beams, especially if the proportion of the permanent load exceeds 75 %; and in many cases part of the imposed load can be considered to be quasi permanent. Insulating board and unreinforced elastomeric sheet can be used as bearing pads. Elastomeric bearings [208] are required when movements between the adjoining components have to be compensated for at the same time as transferring support reactions, i.e. a low-restraint joint must be provided. Rotation and sliding are accommodated by the elastic deformation of the bearing material (see DIN 4141-1 “Structural bearings”). Elastomeric bearings consist of synthetic rubber with a high ageing resistance (trade names: neoprene, Bayprenr). They are available in many forms, unreinforced and reinforced, and are generally covered by National Technical Approval. Elastomeric bearings can handle vertical loads, rotation at the support and structural movements, e.g. due to restraint, although the permissible loads of the relevant approval documentation must of course be taken into account. Thin, unreinforced bearings are adequate when the movements are small. Thicker bearings are required to cope with larger movements, which, however, also cause larger lateral tensile forces if they are unreinforced. Reinforced elastomeric bearings include corrosion-resistant steel plates or textile inlays incorporated during vulcanisation which accommodate the lateral tensile forces within the bearing so that the adjacent parts of the support are subjected to lateral tension only locally, not loaded by the bearing itself.

182


a) Unreinforced elastomeric bearings

The growing popularity of unreinforced elastomeric bearings for buildings and singlestorey sheds is due to their economy and their permanent elastic behaviour. They can accommodate horizontal displacements to a limited extent and minor rotation at the support, and also compensate for some local unevenness. Unreinforced elastomeric bearings are considerably less expensive than reinforced versions and have the advantage that they are not limited to certain forms or types, i.e. bearings can be fabricated to suit particular purposes, even with openings, e.g. for dowels; they are cut to size from large-format sheets. They are being used more and more for the supports to suspended floors, too. Unreinforced elastomeric bearings may only be used with predominantly static loads because there is a risk of creep in the presence of dynamic loads. Generally, elastomeric bearings may be used over a temperature range of s25 to S50 hC. However, their size, positioning and the joint thickness are more important when assessing their behaviour in fire. With a 3 cm thick joint, the rate of burning is J 0.35 mm/ min, which results in minimum dimensions for bearings if they are to satisfy a certain fire resistance rating. Unprotected bearings can be protected against the effects of fire with layers of insulation if their behaviour in fire cannot be assessed. The design of unreinforced elastomeric bearings is dealt with in DIN 4141-15 [214], and [223] contains additional information. The standard deals with bearings whose dimensions comply with the following conditions: a a JtJ J 12 mm bearing thickness: 5 mm J 30 10 bearing plan size: 70 mm J a J 200 mm where a w length of bearing (see Fig. 3.10) The thickness may be reduced to 4 mm if smaller flatness tolerances can be guaranteed (1.5 mm). It is essential to prevent direct contact between the concrete components, even in the case of rotation of the support, and this is the main principle behind specifying the thickness. Only vulcanised products based on chloroprene rubber (CR) may be used for unreinforced elastomeric bearings. Taking into account the permissible local bearing pressure on the adjacent component surfaces, elastomeric bearings may be loaded with an average bearing pressure of sm J 1,2 G S

(52)

183


Fig. 3.10 Arrangement of reinforcement in the region of the beam support (example according to DIN 4141-15)

where the shear modulus G w 1 N/mm2 and the form factor S is Sw

ab ðb J 2aÞ 2ða S bÞ t

(53)

for rectangular bearings, and Sw

D (D w diameter) 4t

(54)

for circular bearings. Drilled holes, e.g. for dowels, may be ignored if they do not exceed 10 % of the bearing area. This therefore results in a compressive stress sm w 10 –12 N/mm2 for standard bearing geometries. But if the specific application conditions given in [216] for column butt joints loaded “purely in compression” apply, then compressive stresses of up to 20 N/mm2 are permissible. In category 2, the lateral tensile force Z due to the prevention of lateral strain in the elastomer must be taken into account by way of Zq w 1,5 F t a 10s5 for a support reaction F. In category 1, the lateral tensile force, in the absence of any more accurate analysis, e.g. through tests, can be determined with the help of the information given in [216]. The reinforcement resisting the lateral tensile force should be positioned as close as possible to the bearing. The tensile splitting force Zs can be calculated according to the relevant publications (e.g. part 2 of Leonhardt’s Vorlesung u¨ber Massivbau). As the calculations represent only rough simplifications, the resulting reinforcement should not be skimped in any way.

184


The reinforcement should be arranged to handle the influences due to both Zq und Zs. The amounts of reinforcement required for splitting tension and lateral tension are then as follows: erf As1 i 1,5 · (0,8 · Zs)/fyd erf As2 i 1,5 · (0,2 Zs S Zq)/fyd where Zs j 0.1 · F In most instances the bottom tension reinforcement provided in the longitudinal direction over the support anyway is adequate for the reinforcement required at a depth of 0.2a for the lateral tensile force. Additional reinforcement is required for the splitting tensile force; the reinforcement in the transverse direction is achieved by reducing the spacing of the shear links. Action effects due to permanent loads parallel to the plane of the bearing (e.g. out-ofplumb, earth pressure, etc.) are not permitted. The following check is relevant for action effects due to restraint and brief external loads for category 1 (but not category 2, where slippage of the bearing is not to be expected or is unimportant): Fx ,Fy w H1 S H2 J 0,05 F

(55)

where H1 external horizontal force H2 restraint force This verifies indirectly that the permissible shear deformation is not exceeded. The angle of rotation a (Fig. 3.11) of a category 1 bearing as a result of elastic and plastic deformations of the components plus the unevenness and skew of the bearing surfaces must satisfy the following condition: zul a J

t 2a

(56)

Where no more accurate verification is provided, a may be determined by adding together the following influences: 1. Probable component deformation under service loads 2. 2/3 of probable component deformations due to shrinkage and creep 3. Skew of 0.01 4. Unevenness of 0.625/a (a in mm) More detailed information regarding the magnitude of the influences given above can be found in [215, 217]. The angle of rotation a should be limited such that there is no direct contact between the concrete components; a minimum distance of 3 mm should be chosen as the limiting value for the point where the concrete components come closest together (Fig. 3.11).

185


Fig. 3.11 Compression and rotation actions on an elastomeric bearing

The eccentricity resulting from the rotation may need to be taken into account for category 1 as follows: ew

a2 a 2t

(57)

when designing the adjoining components. Any influence due to compression of the bearing only needs to be analysed in exceptional circumstances. As the deformation curves are not linear, the compression component due to imposed loads is smaller than the component due to the total load. Extremely smooth contact surfaces have a negative effect on elastomeric bearings because frictional engagement between the different materials is then impossible; release agent and similar substances worsen the situation. So the characteristics of the loadbearing and deformation behaviour require constructional measures that would be unnecessary to this extent when using other bearing materials. Apart from accommodating the lateral tensile forces near the support surface as a result of restricting lateral strain at the contact faces, as mentioned above, protecting the arrises where the permissible bearing pressures are used to the full calls for special care. To do this, the following recommendations should be followed (which are taken from [218]): x

x

Arrises should always be cast with chamfers because this offers the elastomeric bearing only a small contact area in the unreinforced edge zone in the case of excessive swelling. The lateral tension reinforcement (calculated, for example, according to [219]) should not be fixed more than approx. 30 mm below the support surface.

Fig. 3.12 Arrangement of reinforcement below the support for an unreinforced elastomeric bearing (according to [218])

186 x

x

x

x

x


The zone below the bearing shown in Fig. 3.12 must be reinforced. Dimension r1 is chosen to allow the contact area to enlarge without damage upon the bearing being squashed by a vertical load and rotation, in the event of horizontal displacements and in the case of inaccurate installation of the bearing. Reinforcement is to be included according to one of the proposals illustrated. Where the reinforcement simultaneously functions as the tensile bending reinforcement in a corbel, the additional requirement regarding adequate end anchorage of the reinforcement remains unaffected (see section 2.6.2). Concentrations of reinforcement in the vicinity of the front face of the support should be avoided because this weakens the bond between the concrete cover and the loadbearing concrete and can result in shell-like spalling. Care should be exercised to ensure accurate cutting to length, bending and fixing of the edge reinforcement. Bent reinforcing bars from the longitudinal column reinforcement or from corbel bending reinforcement are generally unsuitable as edge protection owing to the large bend radii and the unfavourable distribution transverse to the support. On the other hand, horizontal loops or meshes with closely spaced bars enable an effective and at the same time economic reinforcement layout. Irrespective of the reinforcement near the surface, appropriate tensile spitting reinforcement is always required at an appropriate distance from the surface and in a size and distribution suitable for the splitting tensile forces (Fig. 3.10).

b) Special forms of unreinforced elastomeric bearings

The compression behaviour of elastomeric bearings can be influenced by perforations, studs or other surface textures or cross-sectional forms, also the use of sponge rubber (see below) [220]. The aim is always to achieve a more uniform stress distribution, even in the case of greater unevenness. The voids cause the bearing to yield at first under the load, the degree of yielding decreasing gradually as the voids are filled with the bearing material, so the resistance to the deformation increases progressively. DIN 4114 treats such bearings as normal solid bearings when the thickness t is replaced by the theoretical value tr for a solid pad with the same volume and the same area on plan. Whereas the standard is limited to square, rectangular or circular bearings, a design proposal for strip-type rubber bearings for prestressed hollow-core slabs published in [221] will be described below (Fig. 3.13).

Fig. 3.13 Support for prestressed hollow-core slab (according to [221])


187

A minimum bearing length according to section 3.1.2 is required for these strips, but this value must be equal to 1/125 x span at least. There must be a clearance of 30 mm between the edge of the bearing and the edge of the concrete in order to avoid spalling at the edges. This therefore results in a bearing length of approx. 40 mm when we take into account the aforementioned minimum bearing length. Tests on prestressed hollow-core slabs with typical dimensions and spans bearing on rubber strips with Shore hardnesses 40 and 60 and sponge rubber with a density of 0.5 g/cm3 have resulted in a recommendation for the latter of 20 mm wide strips with thickness t w 8–10 mm. This means that a compression of 3– 4 mm as a result of the load and compensating for unevenness leads to a residual gap under full load of min. 2–3 mm. The sponge rubber recommended for bearings is supplied in rolls and chloroprene rubber is also available. c) Sliding bearings

Unreinforced or reinforced elastomeric bearings of suitable thickness can be used where small relative movements between two components must be permitted. But for larger movements it may be necessary to install special sliding bearings. There are many types available but they do not have National Technical Approvals, instead in some cases are subject to voluntary official quality control measures. These bearings for buildings consist of lubricated or non-lubricated films (0.2– 0.5 mm) or sheets (3–5 mm). The materials used are polyethylene (PE), polypropylene (PP), polyvinyl chloride (PVC), polyamides (PA) or polytetrafluoroethylene (PTFE), with the last of these being the most suitable, but also the most expensive (trade names: Teflon, Hostaflon, etc.). In the meantime, carbon fibre-reinforced plastics (CFRP) are also being used. The films and sheets are enclosed in foams or elastomers because otherwise they are too thin to compensate for unevenness or to prevent excessive edge pressures in the case of rotation. Only sliding bearings adequately laminated to an elastomer to achieve a minimum thickness of 4 mm in total may be used as intermediate pads between precast concrete components. This is not a true sliding bearing, but rather a “deformable sliding bearing”. The coefficient of friction depends on pressure, material, temperature, lubrication, sliding velocity, bonding at perimeter and number of movements, i.e. a whole range of parameters. The friction coefficients determined by the manufacturers – often under laboratory conditions only – differ from those suitable in practice, in some cases quite significantly. A value of m w 0.10 can be assumed as a cautious characteristic value [222]. d) Reinforced elastomeric bearings

Reinforced elastomeric bearings for higher loads are square, rectangular or circular on plan. Their reinforcement is in the form of flat sheet steel or textile plies positioned at a uniform spacing and symmetrically about the central plane. They are bonded to the elastomeric plies by way of hot vulcanising. However, before opting for a more complex, more expensive reinforced elastomeric bearing, it is always important to check whether the requirements to be satisfied by the bearing can be met by an unreinforced elastomeric bearing. This is the case for the majority of bearing details in precast concrete construction.

188


The use of reinforced elastomeric bearings is covered by DIN 4141-14 “Reinforced elastomeric bearings” and National Technical Approvals. Ref. [208] contains further details of reinforced elastomeric bearings. Reinforced elastomeric bearings differ from unreinforced elastomeric bearings in the following ways: x

x

x

x

x

Whereas unreinforced elastomeric bearings are made from sheets that can be cut to suit the respective application, reinforced elastomeric bearings are only available in the sizes fabricated by the suppliers. The permissible pressures depend on the size of the bearing and lie between 10 N/mm2 for the small ones and 15 N/mm2 for the large ones. The reason for the higher values of the reinforced bearings, compared to the unreinforced bearings, is that squeezing-out of the bearing as it prevents lateral strain is checked by the sheet metal incorporated during vulcanisation. The permissible shear deformations, related to the ply thickness t, are the same as for unreinforced elastomeric bearings. But as thicker bearings with T w n · t are permissible for reinforced bearings, the permissible shear deformations are increased by a factor of n for the same value of t. In terms of rotation, the total rotation 4 w n · a applies accordingly, where a is the angle of rotation per elastomer ply. Apart from the tensile spitting forces caused by local loads, it is not necessary to take into account any further lateral tensile forces for reinforced elastomeric bearings.

3.1.4

Elastomeric bearings to DIN EN 1337

Elastomeric bearings to DIN 4141 were described in the previous section. This standard has in the meantime been withdrawn and replaced by the new bearings standard DIN EN 1337. The new standard consist of 11 parts, and part 3 deals with elastomeric bearings (see also [213]). As elastomeric bearings were further developed and achieved permissible bearing pressures i 12 N/mm2 and exhibited a loadbearing behaviour that increasingly deviated from the provisions of DIN 4141, it became clear that a new standard was required. A new bearings standard was drawn up over several years in the course of the European harmonisation of the new generation of standards. Furthermore, the many and various technical developments and the diverse requirements placed on bearings have led to the following regulations: a) There will be no bearing categories in the future. The edition of DIN EN 1337-3 “Structural bearings – Part 3: Elastomeric bearings” currently valid (Jul 2005) covers unreinforced elastomeric bearings made from CR (chloroprene rubber) and NR (natural rubber) for relatively low vertical loads (up to approx. 8 N/mm2) and predominantly static actions. However, the introductory decree excluded natural rubber, which means that currently only elastomeric bearings made from chloroprene rubber are covered. As a result of the limited loads, the analysis for determining the bearing pressures for these bearings has been much simplified in the standard.


sEd;m w

Fzd J 1,4 Gd S J 7 Gd A

189

(58)

where Fzd design value for vertical load A plan area of bearing Gd design shear modulus of elastomer S form factor for elastomeric material The bearing forces and bearing movements are determined as characteristic values according to the rare loading combination of DIN 1055-100. The characteristic values resulting from the individual actions (increased by the respective partial safety factors) are used for calculating the resulting design values for movement – displacements and rotations – and the forces at the ultimate limit state. b) All bearings that cannot be designed according to a) require some form of National Technical Approval. This is the case, for example, when bearings according to a) are to be used for higher loads, or other types of bearing or other materials (EPDM, ethylene polypropylene diene monomer rubber) are to be used. Besides some form of approval, these bearings require a design concept. At the moment, a so-called application standard is being drawn up which will specify a method of analysis. Therefore, all the information given below reflects the current status of discussions in the committee responsible for the new standard and has not been finalised. The application standard must function within the scope of two limiting values: the lower value is formed by the simple verification to DIN EN 1337-3 and the upper value is the current limit of a 20 N/mm2 bearing pressure based on experience. In terms of the climatic influences, we distinguish between components fitted within insulated building envelopes and components exposed either frequently or permanently to the outside air. The analysis of the bearing is carried out on the basis of a deformation-related calculation concept instead of DIN EN 1337-3. In doing so, we distinguish between an accurate and a simplified analysis. The average bearing pressure is limited in order to restrict the lateral spread or sagging of the bearing: sEd,m w

Fzd J sRd,m A

The design value of the admissible average bearing pressure sRd,m can be found in the relevant National Technical Approval or DIN EN 1337-3. Both the angle of rotation of the component at the support and a geometrical imperfection should be taken into account for the rotation at the support aEd,tot : aEd,tot w aEd,component S aimp

190


The angle of rotation aEd,tot of the bearing should be limited in such a way that there is no contact between the edges under the design values. As the bearings themselves can accommodate very large pressures or deformations without the elastomer being damaged, avoiding damage to the adjoining components constitutes the actual limit state. The eccentricity as a result of the restoring moment due to rotation of the bearing must be taken into account when designing the adjoining components. In future this data will be available in the relevant National Technical Approval. The shear distortion in the bearing due to component displacements and brief external loads tan gEd should be limited as follows: tan gEd J tan gRd J 1,0 In addition, it is also necessary to check that non-anchored bearings cannot slide. The application standard and the future National Technical Approvals for elastomeric bearings for buildings have been coordinated with each other. 3.2

Tension joints

3.2.1

Welded joints

These days, only weldable structural steels are appearing on the market and the steel grades for structural steelwork approved to DIN 18800 -1 plus pipe steels for seamless and welded hollow sections are also all suitable for welding. This means that the permanent loadbearing connections for precast concrete buildings are frequently in the form of welded joints. The design, fabrication and quality control of welded joints in reinforced concrete construction is covered by DIN 1045-1 and DIN EN ISO 17660 parts 1 and 2 “Welding of reinforcing steel”; part 1 covers loadbearing welded joints, part 2 non-loadbearing joints, i.e. those required for transport and/or erection only. One example of the latter is the standard meshes, which are fabricated with plain overlapping joints. Only loadbearing welded joints are considered here. All welded joints (and this applies to non-loadbearing joints as well) may only be produced by welders qualified in accordance with ISO 9606-1 (fillet weld test) and who have completed additional training covering the welding of steel reinforcing bars. Furthermore, the contractor or fabricator must employ a supervisory person according to ISO 14731 who has specific knowledge of the welding of steel reinforcement. DIN 1045-1 Table 12 specifies the permissible welding methods. The method of welding used almost exclusively in precast concrete construction, apart from stud welding for shear connectors, is electric arc welding. We distinguish here between the manual electric arc welding often used (E), with coated electrodes, and shielded arc welding, often referred to as metal active gas (MAG) welding. The latter is particularly suitable for factory

3.2 Tension joints

191

fabrication, and may only be carried out in the open air when a protective tent is erected around the work to prevent the shielding gas being blown away by the wind. Also possible these days are welded connections involving stainless steel grades 1.4401 and 1.4571 to DIN EN ISO 10088 (see also DIBt Approval Z-30.3- 6 covering products, connections and components made from stainless steels) and carried out according to DIN EN ISO 17660 -1. However, special electrodes are required for this. When designing joints involving both normal and stainless steel, the different thermal expansion behaviour of these two types of steel must be taken into account in order to avoid restraint stresses and failure. Galvanic corrosion can occur when these two types of steel are in contact in the presence of an electrolyte, e.g. water, which, however, is not a problem when the steel is embedded in concrete. DIN EN ISO 17660 -1 provides detailed information about the design and detailing of welded connections between steel reinforcing bars and between reinforcement and steel sections. The basic rule is to provide fillet welds (Fig. 3.14) and lapped or spliced joints (Fig. 3.15), which should be preferred to butt joints or T-joints (Fig. 3.16) because they are easier to produce and have significant loadbearing reserves.

Fig. 3.14

Fillet welds along the sides of reinforcing bars or other steel parts to DIN EN ISO 17660-1

Fig. 3.15 Lapped and spliced connections between reinforcing bars to DIN EN ISO 17660-1 for loadbearing joints

192


Fig. 3.16 Fillet welds at ends of reinforcing bars to DIN EN ISO 17660-1

Fillet welds around the ends of bars (T-joints) are very prone to errors. The recommendation for structural calculations (when there are no results available from welding tests) is to carry out such fillet welds only for those details where the bar passes through a plate (Figs 3.16a and 3.17) with exactly the same load-carrying capacity as the bar. This approach is based on company tests and complies with [227]. Where the bar must finish flush with an anchor plate for constructional reasons, e.g. at short supports or corbels, then the type of joint shown in Fig. 3.16c is not to be recommended. As the aforementioned tests have revealed, such a joint carries I 50 % of the force in the bar in some situations. The type of welded joint shown in Fig. 3.16b can be used if a flush fitting is unavoidable, but this detail can carry only 75 % of the load in the bar, unless specific suitability tests have proved that the full load-carrying capacity can be assumed. The loadbearing effect of the welded joint shown in Fig. 3.16a is the result of the wedging effect of the relatively small annular weld in particular. Failure occurs through a coneshaped shearing-off of the projecting bar, which is why this should not be less than 1.0 · ds if at all possible. However, the aforementioned tests have shown that it is not necessary to provide a weld with 0.4 · ds. Instead, a simple wedging with a throat size of a w 7 mm is adequate (Fig. 3.17). Where the loads are not predominantly static, DIN 1045-1 Table 12 specifies that butt joints between bars in tension may only be carried out with flash welding (FW). Electric arc and MAG welding are only suitable for bars in compression, and then only within certain limits.

193

3.2 Tension joints

Fig. 3.17 Welded end anchorage with circular anchor plate (according to [157])

Welded joints in structures in seismic regions must be given special attention. DIN 4149 section 8.3.5.3 specifies that welded connections between reinforcing bars are not permissible for any stability components for which a higher degree of ductility is presumed (class 2). The requirements to be met by filler metals given in DIN 4149 section 9.3.1.3 must be adhered to for welded connections in steelwork. The welded connections in precast concrete construction are frequently based on reinforcing bars projecting from the precast concrete components which are then either welded directly together or to splice plates and the whole joint then encased in concrete on site (see Fig. 2.44). The projecting bars must be long enough to permit minimal bending in order to accommodate tolerances. Good access for the welder and the welding equipment is essential. The reader should consult FDB leaflet No. 2 (available in German only) for information about corrosion protection. 3.2.2

Anchoring steel plates, dowels, studs and cast-in channels

Fixing technology in reinforced concrete construction has evolved into a separate field in recent years. This technology includes cast-in parts, e.g. channels, rails, plates with welded headed studs or pigtail anchors made from ribbed bars with threaded couplers pressed on, and site fasteners in the form of expansion, undercut or bonded anchors (Figs 3.19 and 3.20). This subject has its own chapter in the Beton-Kalender [230, 238]. It will therefore not be covered in any detail here. These cast-in parts are being increasingly regulated by way of European technical specifications [228], which enable the manufacturers to create specific design software and thus simplify the use of their fixings. For example, the so-called CCD method (CCD w Concrete Capacity Design) represents the current standard for designing anchors and anchor plates. There is a similar development for cast-in channels and rails [229]. A compilation of typical anchor channels and rails plus typical fasteners for precast concrete components and façade elements can be found in the Beton-Kalender [231, 238].

194

Fig. 3.18


Welded joint for a suspended floor slab

Fig. 3.19 Examples of drilled anchors: (a) force-controlled expanding anchor, (b) undercut anchor, (c) bonded anchor

Fig. 3.20 Examples of cast-in parts: (a) channel, (b) plates with welded headed studs, (c) pigtail anchor

3.2 Tension joints

3.2.3

195

Shear dowels

Shear dowels are frequently used in precast concrete construction in order to secure the positions of precast concrete components that are in direct contact. The shear dowel intersects the contact face between the two parts and must resist any shear forces occurring in the joint. Shear dowels are dealt with in detail in [232–234, 239]. The design criteria are the compressive stress in the concrete at the point of fixity of the dowel and the bending of the dowel. Whereas the bending moment in the dowel can be calculated relatively accurately, determining the load-carrying capacity of the concrete, e.g. with the help of the subgrade modulus, is a problem. Shear dowel joints can fail (Fig. 3.21) by way of – high bearing pressure on the concrete (a), – cracking of the cross-section (b and c), and – excessive bending in the dowel. As both failure of the steel and failure of the concrete are possible, both of the following analyses (taken from [232]) should always be carried out. The shear force that can be accommodated by the dowel is calculated taking into account the plasticity reserves in the steel with a factor of 1.25 as follows: Fu w 1,25

fyk W (a S xe )

where fyk yield stress of dowel steel W moment of resistance of dowel a lever arm of force xe theoretical fixity depth for dowel A safety factor of g w 1.75 is adequate for failure of the steel. Considering the risk of local spalling, selecting xe w d w dowel diameter is to be recommended. The embedment length required lies between 5d and 6d; using 6d every time is sensible.

Fig. 3.21 Failure mechanisms in shear tests [232]: (a) local spalling, u? =d I 8) (b) inadequate edge distance (€ujj =d I 8), (c) inadequate edge distance (€

196


The failure load of the concrete is as follows: Fu w 0,9

fck (d2,1 ) ½kN (333 S a 12,2)

where d dowel diameter a lever arm (in mm) of load application to be used fck cylinder compressive strength of concrete to DIN 1045-1 (in N/mm2) A safety factor of g w 3 is recommended. A formula according to Rasmussen given in [232] leads to roughly the same result assuming a w 0 for dowel diameters d w 16–25 mm: qffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi fck fyk Fu w 1,3 d2 A safety factor of g w 5 is recommended in this case. Where failure of the concrete beneath the point where the dowel exits the component is prevented, e.g. by a steel disc with a diameter of min. 7d welded to the dowel or by a transverse compressive force acting on the joint, then the permissible load on the concrete may be roughly doubled. A prerequisite for the above equations is an adequate minimum edge distance of u¨|| and u¨^ j 8d. For constructional reasons alone, smaller edge distances should not be used with unreinforced concrete. The concrete can be strengthened with reinforcing bars in the case of smaller concrete dimensions and larger dowel shear forces. The cross-sectional area of the reinforcement can be calculated using the following equation: As w

1 FEd c fyd

(59)

where c is taken from Fig. 3.22. Mesh reinforcement with a bar spacing of 50 mm and a bar diameter J 8 mm has proved worthwhile. In such a mesh, up to five bars parallel to the load may be considered as effective. Alternatively, loops in double shear with a bar diameter J 12 mm may be placed around the dowel and anchored in the opposite direction to the force (Fig. 3.23). To conclude this section, information about two interesting proprietary shear dowels, which are marketed as “shear load connectors” and permit movement in the longitudinal direction of the dowel (Fig. 3.24). Type I is designed to be fitted through the mould. The steel bar has a bituminous coating over half its length which prevents a bond with the concrete and therefore permits movement in the longitudinal direction provided there is a space or compressible foam material at the end of the dowel.

197

3.2 Tension joints

Fig. 3.22

Values for the factor in eq. (59)

Fig. 3.23 Effective forms of reinforcement for shear dowels [232]

Fig. 3.24

Sliding dowels (Speba system)

198

Fig. 3.25


Types of shear connector system for different degrees of longitudinal displacement

Type II includes a sleeve for the dowel which is nailed to the inside of the mould; there is no need to drill a hole in the mould for the dowel. After demoulding, the dowel is inserted into the sleeve and cast into the adjoining component once the joint is finished. Such shear connectors are highly advisable when, for example, movement in the longitudinal direction of the connector is necessary once a bearing or expansion joint has been compressed and a shear force still has to be resisted. The development of these proprietary “shear load connectors” has continued in the meantime and now they are available in the forms shown in Fig. 3.25, available with different degrees of freedom and covered by National Technical Approvals [231, 240]. In conjunction with suitable additional reinforcement, these fittings can be used as loadbearing connections between components. 3.2.4

Screw couplers

A whole range of screw couplers is available for connecting reinforcing bars – for transferring tensile forces in particular. However, most of these products are unsuitable for connections between precast concrete components because they cannot compensate for any tolerances in the longitudinal and transverse directions, or at best are only able to cope with minimal tolerances, and also because the couplers themselves or the equipment needed to install them require too much space. Such screw couplers are, however, very useful for connecting loose reinforcing bars laid in grouted joints or passing through sleeves, or reinforcing bars that project from precast concrete components (so-called starter bars) and are cast into in situ concrete. Detailed information on the screw couplers described below can be found in [231]. Screw couplers have become very popular in precast concrete construction over recent years. The problem of the fitting accuracy of the two precast concrete components with respect to the screwed connection has been alleviated by using a system with fixed, cast-in bolts, normally with a screw coupler to avoid having to penetrate the mould, and a cast-in part with appropriate tolerance options. The permissible tolerances lie between 3 and 8 mm, depending on the size of the cast-in part. Using this system it is now possible to design column, beam and wall joints for both loadbearing and non-loadbearing details (see also Figs 2.63 to 2.68).

3.2 Tension joints

3.2.5

199

Transport fixings

Precast concrete components are not produced in their final position and so how and at which points they can be lifted, turned and transported must be considered at the design stage. The type, positioning and size of the fixings must be planned, also the lifting equipment. Availability must be checked in the case of the latter. As transport fixings do not have any effect on the safety and stability of the finished structure, they are often treated perfunctorily during the planning work. The engineer performing the structural analysis normally does not consider himself or herself to be responsible for the transport fixings; this is because in Germany the employers’ liability insurance associations (Berufsgenossenschaften) are responsible for safety during transport. However, these associations are only interested in the quality of the crane slings and hooks, which means that according to the building authorities nobody is actually responsible for the transport fixings so critical to the safety of the workers on site! The responsibility is therefore entirely down to the manufacturer of the precast concrete components. Only tried-and-tested transport fixing systems should be used for larger loads. Ref. [231] shows some of the many transport fixings available. According to the safety rules of the employers’ liability insurance associations, transport fixings must be designed to carry three times the nominal working load. The latter is specified for a concrete strength of 15 N/mm2, which must be available at the time of lifting [133, 235]. In the meantime, a study group has been formed by VDI Bautechnik and the Studiengemeinschaft fu¨r Fertigteilbau e.V. in order to draw up detailed rules for the design and use of transport fixing systems and to prepare a VDI directive on this subject [241]. The following points must be considered when sizing transport fixings: x

x

x

x

Adhesion forces of considerable magnitude can occur during demoulding. They can reach figures of 1 kN/m2 in the case of steel moulds treated with release agent, or 3 kN/m2 in the case of rough-sawn timber moulds. Indeed, when demoulding double-T floor units cast in older rigid moulds, values equal to twice the self-weight have been measured. When fixings are subjected to diagonal pull, higher forces in the ropes (Fig. 3.26a) and additional bending stresses in the fixing sleeves must be taken into account. Angles i 60h are not permitted. Diagonal pull can be avoided by using a spreader beam. Special diagonal pull loops may have to be used (Fig. 3.26b). Only two loadbearing fixings and ropes may be assumed in the design for a statically indeterminate suspension system. The designer is therefore recommended to set up a statically determinate loadbearing system by using a spreader beam or rocker attachment that apportions approximately identical loads to all fixings (Fig. 3.27). If tilt-up moulds are not available for wall panels and columns cast horizontally, the edges of the concrete components are damaged again and again when attaching slings and raising the units because of the inadequate edge distances of the transport fixings. The fixings must be suitably rigid and secured with reinforcement (Fig. 3.28). While

200


Fig. 3.26 (a) Higher rope forces with diagonal pull, (b) cable loops for diagonal pull (Schro¨der)

x

raising the units, only half the load acts on the transport fixings because the other half is still supported on the factory floor. Non-uniform and impact loads can occur when lifting a precast concrete component out of the mould, also during raising and lowering them either in the factory or on the building site. The magnitudes of those loads depend on the skill of the crane operator. Cranes with precision hosting mechanisms are available in the precasting plant and these are generally operated by experienced personnel, which means that the im-

Fig. 3.27

Statically determinate lifting method

3.2 Tension joints

Fig. 3.28

201

Transport fixings for a wall panel cast horizontally

pact loads remain low and can generally be ignored even though the concrete strength is at its lowest during the handling in the plant. This is also true for the latest mobile cranes. A global increase in the safety factor for the transport fixings system in order to cover the impact events that, theoretically, can occur during improper handling is not justifiable on economic grounds. It is up to the respective personnel on site and in the factory to handle the components carefully. Transport fixings that can be detached with a cable or pneumatically via remote control help to ease positioning operations in many instances and reduce the risks to erection personnel. Neglecting the duty of care necessary during the planning of transport fixings counts is not worthwhile when we consider what accidents and damage might be caused by falling parts and also the time wasted when attaching crane slings and hooks. The only important aspect is that the cast-in parts, of which there are normally two or four per component, should not be too expensive (because they remain in the component and are probably never used again). 3.2.6

Retrofitted corbels

Where corbels have to be fitted afterwards because of the type of fabrication employed, e.g. walls built using climbing formwork or slipforming methods, then it is possible to employ the options shown in section 2.6.2, with butt-jointed reinforcing bars and shear key joints. And even where there are no connectors present in the structure, it is still possible to attach corbels subsequently by exploiting friction and dowel actions to transfer the forces. Tensile forces must be transferred by threaded fixings. Fig. 3.29 shows two options for retrofitting a corbel, both of which require bolts. Yielding of the connected parts as a result of small inaccuracies and bolt relaxation can be avoided by pretensioning the bolts with a tensile force Z calculated beforehand. The pretensioning can be applied with hydraulic presses like those used in prestressed concrete construction, but using a torque wrench is more convenient, which, however, also induces torsion in the bolt as well as tension.

202


Given:

V, e, z = 0.8 h corbel width b concrete grade (of weaker part) Wanted: pretension Z or tie cross-section Stress in concrete must be checked Pretension: e/z ≤ 1.23 : Z = 2.15 V e/z > 1.23 : Z = 1.75

–ez

Tie cross-section: As = Z/perm σe (pretension tie to Z)

V, e, z, d0, t concrete grade Wanted: pretension Z or tie cross-section Stress in concrete must be checked Given:

Tension: Z = –ez V Tie cross-section: As = Z/perm σe (pretension tie to Z) Concrete stress:

Concrete stress: recommended for dowel: d ~ – t [237] where tan α = z/e → sin2 α [236] Reinforced concrete corbel attached with pretensioned HSFG bolts

Fig. 3.29

Steel corbel with dowel

Retrofitted corbels

The pretensioning force Z that can be achieved with a torque MD and lightly oiled, highstrength friction-grip (HSFG) bolts can be calculated with good accuracy using the following equation: ZðkNÞ w

5 MD ds

(60)

(MD in Nm and bolt diameter dS in mm) The following equations can be used to estimate the dimensions of the anchor plate for transferring the tensile force in the bolt to the concrete: pffiffiffi plate thickness: t reqd [mm] w 3.4 3 Z (Z in kN) plate area: AD reqd [cm2] w 0.8 · Z (Z in kN) These are based on tests [236] carried out with concrete grade C20/25 (B 25) and a through-hole with a diameter of about 1.5ds. The reader is referred to the very detailed information given in [236] regarding the fire resistance and corrosion protection of bolted connections. All components that are to be retrofitted with corbels must of course be checked to ensure that they can carry the additional loads. Fig. 3.29a shows a reinforced concrete corbel fixed to a concrete component with pretensioned HSFG bolts. The friction force needed to carry force V is activated by tensioning

203

3.3 Shear joints

with force Z, which is why force Z for this corbel detail has to be much greater than that in Fig. 3.29b. The authors of [236] provide further information about this corbel detail: x

x

Filling the joint with grout is unnecessary, and this does not mean that extreme requirements are placed on the flatness of the contact faces. When the HSFG bolts have been properly galvanised, then the drilled hole for each bolt does not have to be pressure grouted in order to reduce the risk of corrosion. This simplifies the exact adjustment of the corbel. An appropriate final detail – mortar fill, possibly in recesses with special mortar, or an “Isoternit” protective cap – can guarantee a fire resistance rating of F 90 to F 120.

Tests [237] on the corbel detail shown in Fig. 3.29b have revealed that this type of corbel is ideal for retrofitting situations, where a corbel was never envisaged in the first place. The sizing is carried out according to the rules of structural steelwork. The hole for the round dowel is drilled with a core drill. The hole should be only a few millimetres larger than the diameter of the dowel that is grouted into the hole. In the tests, this type of corbel was loaded to failure with 2.1 times the service loads. 3.3

Shear joints

3.3.1

General

The design of shear joints for cross-sections subsequently finished with in situ concrete has already been dealt with in section 2.6.1. According to DIN 1045-1, the same equation can be used for other types of shear joint as well. Fig. 3.30 shows in graphical form the loading components that contribute to transferring shear. The first component comprises the adhesion between the grout in the joint and the precast concrete component, the second component corresponds to the friction force due to an axial stress perpendicular to the joint, and the third component is the reinforcement, which also activates a friction force and can be explained by way of shear-friction theory [242].

Fig. 3.30

Components contributing to shear transfer

204


Fig. 3.31

Shear-friction theory

In shear-friction theory (see Fig. 3.31) it is assumed that even a minor crack at a joint subjected to shear is sufficient to transfer the load to the steel crossing the joint. This happens because with a relative displacement of the contact faces the roughness in the crack separates the faces and therefore the steel bars across the joint are subjected to a strain. This gives rise to a compressive force in the joint which enables the shear force to be resisted by friction forces. In principle, the reinforcement therefore performs the same function as an external compressive force applied perpendicular to the axis of the joint. The dowel effect of the reinforcement is small by comparison (see section 3.2.3) and is generally neglected. If the reinforcement crosses the joint at an angle, the proportion projected onto the shear joint contributes directly to resisting the shear force acting in this direction. 3.3.2

Floor diaphragms and wall plates – in-plane shear forces

(see also sections 2.2.5 and 2.2.6) According to DIN 1045-1 section 13.4.4, a suspended floor made up of precast concrete elements can be regarded as a loadbearing plate provided it forms a coherent planar surface in its final condition, the individual elements of the floor are interconnected with compression-resistant joints and the loads in the plane of the plate can be resisted by arch or truss action together with the perimeter members and ties reinforced for this purpose. The ties required to achieve the truss action can be formed by reinforcing bars laid in the joints between the precast concrete elements and anchored accordingly in the perimeter members. The reinforcement in the perimeter members and ties must be verified by calculation. However, various truss action models with different strut angles are conceivable (Fig. 3.32). At first sight, forming ties in every joint between precast concrete components when assuming steep struts would seem to call for more reinforcement. However, with a concentrated diagonal tie, it must also be possible to install the reinforcement required within the joint. Furthermore, there is no need to verify the joints because they are not intersected by the struts.

205

3.3 Shear joints

Fig. 3.32 Two truss action systems with diagonals at different angles

The edges of the longitudinal joints of floor diaphragms made from aerated concrete components or prestressed hollow-core units cast using slipformers or extruders are relatively smooth because of the production process. Various research projects have therefore been carried out for these products and these have verified that the stiffness of perimeter or internal connecting beams, which act as dowels, make a considerable contribution to transferring shear forces. A simple design concept for one- or two-storey residential buildings has been developed for aerated concrete components based on full-scale tests [247]. DIN 1045-1 section 10.3.6 can be used to verify the shear joints in prestressed concrete suspended floors [248]. In doing so, it should be noted that the joints must be classed as smooth and the value for the shear force may not exceed hF · 0.15 N/mm2. The effective shear force may be distributed over the entire length of the joint as follows: vEd w

A J vRd,ct J hF 0,15 N=mm2 L

where hF w h - 20 mm effective joint depth. Only predominately static loads are permissible. Whereas DIN 1045-1 section 13.4.4 assumes floor diaphragms, these days walls, made up of precast concrete components, are to an increasing extent also being produced to function as loadbearing plates in various structural systems. The analyses can be carried out according to DIN 1045-1 section 10.3.6. Up until now, the curves shown in Fig. 3.35, derived by Schwing [67] from comprehensive tests, were the most widely used for precast concrete construction. These allow various shear key geometries for various loading situations, as depicted in Fig. 3.33, to be calculated and designed. Although the curves should also apply to smooth joints, with a somewhat higher factor of safety, a shear key is always recommended for loadbearing joints. Taking into account the global factor of safety of 2.5 recommended by Schwing for a shear key joint, this results in the following shear capacity per unit length for a B/Fu ratio J 0.5:

206


Fig. 3.33 Typical loading conditions in the joints of precast concrete plates [67]

Fig. 3.35

Fig. 3.34

Shear key geometry

Curves for determining the amount of reinforcement required in the joint [67]

207

3.3 Shear joints

vRd w

b1 k glc

rffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi fck B a S b r fyk SsN j vEd Fu

(61)

where glc w 1,76 B h1 b1 w Fu h2 b a w 0,04 [MN/m2] b w 0,44 [s] r w amount of reinforcement

Table 3.1

Factor k for eq. (61) k

k

C12/15

0.95

C30/37

0.908

C16/20

0.95

C35/45

0.885

C20/25

0.95

C40/50

0.862

C25/30

0.93

C45/55

0.839

fyd w fyk=g [MN=m2 ] s sN w compressive stress

0 J r fyk S sN J 3,80 [MN=m2 ] k Using the shear key geometry given in DIN 1045-1 Fig. 35 results in a B/Fu value of 0.5. Fig. 3.36 compares the curves according to Schwing with the DIN 1045-1/A1 analysis. Excellent agreement has been achieved for typical axial forces. In practice the reinforcement required is often concentrated in the transverse joints crossing the shear joint (see Fig. 2.59). To allow for this, Schwing recommends increasing the amount of reinforcement by a factor of 1/0.85. A number of manufacturers have developed special cast-in fittings for joints to simplify the provision of shear key and reinforcement. These form, or rather replace, the shear key with profiled sheet metal, and the reinforcement is provided in the form of looped

Fig. 3.36

Comparing curves with DIN 1045-1/A1 for a joint geometry of B/Fu w 0.5

208


Fig. 3.37 Shear connectors made from profiled sheet metal fittings (Pfeifer system)

cables to simplify installation (Fig. 3.37). Such fittings can accommodate shear forces of up to vRd w 90 kN/m. After installing these units, the joints are finished with highstrength mortar, which is specified in the appropriate approval documentation along with the permissible tolerances and the minimum component thicknesses. Interaction with out-of-plane forces (e.g. wind loads on a wall plate) is possible. It should be noted that as a result of the cable loops, cracks are always somewhat wider (Dw w 0.1 mm) than is the case with reinforcing bars. Special shear connectors have been developed for connecting wall plates and columns (Fig. 3.38).

Fig. 3.38 Shear connector for wall junctions (Pfeifer system)

3.3 Shear joints

3.3.3

209

Joints in suspended floor slabs – out-of-plane shear forces

The joints in suspended floors made up of precast concrete components must also transfer shear forces perpendicular to the plane of the floor as well as those in the plane of the floor, i.e. acting in the longitudinal direction of the joints. Fig. 3.39 illustrates failure mechanisms for various reinforced joints. DIN 1045-1 section 13.4 contains standard details for this, which enable joints to transfer shear forces by way of – a concrete filling with or without transverse reinforcement (Fig. 3.39), – welded or dowel-type connections (Fig. 3.40), or – a reinforced concrete topping. A reinforced concrete topping is always required for the transverse distribution where the loads are not predominantly static. Comprehensive tests with slab thicknesses of 10 –20 cm and various joint forms are described in [244] and [245]. Unfortunately, the design proposals derived from these tests for unreinforced joints contain a mistake, which, however, has since been corrected [246]. Fig. 3.41 shows options for the design of reinforced joints. The parameters have been adjusted to suit the new edition of DIN 1045-1. DAfStb booklet 525 [147] specifies the admissible shear force for unreinforced joints, which is based on [245], as follows: rffiffiffiffiffiffiffiffiffiffiffiffiffi 1,44 h 3 fck,cube VR,joint,perm w VR,joint,0 10 45 where VR,joint,0 w 7.5 kN/m

Fig. 3.39 Failure mechanisms for unreinforced, dowel-reinforced and loop-reinforced joints (according to [244] and [245])

Fig. 3.40

Examples of joints between precast concrete elements to DIN 1045-1 (dims. in mm)

210


Fig. 3.41 Transverse forces in reinforced joints (according to [244] and [245])

However, this equation only applies for concrete grades up to C45/55 and slabs up to 20 cm deep. The joints in deeper slabs should be similarly narrow, i.e. altered in proportion to their depth only. In unreinforced joints it is the concrete nib that fails. We can therefore assume that the admissible joint shear force will then increase roughly in proportion to the depth (of the concrete nib). The ratio of expansion force to shear force is smaller anyway, corresponding to the steeper angle of the compressive force. The authors’ own tests on 10 cm deep slabs with unreinforced joints with a form as shown in Fig. 3.42 and a C20/25 filling to the joint resulted in minimum shear forces at failure of Vu w 27.2 kN/m, which with a safety factor of 3.0 corresponds to a permissible shear force of approx. 9 kN/m. The horizontal component H of the inclined compressive force D, which in this case is equal to roughly twice the shear force V, acts as an expansion force and must be transferred via the floor diaphragm to the longitudinal reinforcement in the

3.3 Shear joints

211

Fig. 3.42 Joint between double-T floor elements, Zu¨blin 6M system

transverse joints (Fig. 3.42). In reinforced joints the joint expansion forces are resisted by the joint reinforcement itself. Regarding the joint form, it can generally be said that a nib according to Figs 3.40 and 3.42, where the depth of both upper and lower nib is about 1/3 h, is the best solution. The joint should be kept as narrow as possible. At the bottom it only has to compensate for the tolerances and at the top it should be just wide enough to allow the mortar filling to be installed and compacted properly and also so there is enough space for any joint reinforcement, including any laps. Reinforcement in the concrete nib is recommended where the shear forces are significant, but is unnecessary in the case of low loads (e.g. wind loads on walls). This is because the tensile strength of the concrete in the nib is critical and the admissible shear force has already been multiplied by the appropriate safety factor. Ref. [249] describes tests on joints in hollow-core slabs. These floor systems were designed for loads that are not predominantly static, for uniformly distributed loads of up to q J 12.5 kN/m2 and fork-lift trucks up to 35 kN. The analysis of the transfer of shear forces transverse to the joint may be carried out according to [147], for prestressed hollow-core slabs as well, but only for imposed loads of q I 2.75 kN/m2 and with an upper limit according to the approval documentation.

4

Factory production

4.1

Production methods

The factory production methods for precast concrete have continued to develop further towards industrialised, i.e. mechanised, methods in recent years. And at the moment, automation, too, with the help of the latest CAD/CAM technologies, is also starting to infiltrate precast concrete construction. The precast concrete industry is being forced to invest substantial sums of money to secure its share of the market in the face of competition from other forms of construction. In this respect, it is the flexibility of the production facilities that is attracting great attention because large repetitive series are becoming things of the past [250]. The majority of industrialised methods used for the production of structural precast concrete components for buildings can be assigned to one of the following two basic techniques: – Circulation flow production – Production in long lines or beds Both methods require a certain throughput, however. The circulation flow production method [251], in which the elements, on pallets, are transported through the factory from one operation to another on roller conveyors or traversers, is the typical method for planar elements such as wall panels and floor units. These days, flow production systems are designed to achieve good flexibility. Flow production has two main advantages: Better organisation of the entire production procedure. The materials required can be made available without internal transportation and the individual workers carry out the same work at the same place each time. x Lower plant costs because the individual operations are carried out at workstations specially designed for those operations and, for example, vibratory compactors or hydraulic systems for moulds only need to be provided once and therefore can be equipped with more functions. x

Besides the very common horizontal flow method with longitudinal conveyors and transversers for heat treatment in curing chambers, there is also space-saving vertical flow method with longitudinal conveyors on upper and lower levels connected via raising and lowering stations. The actual production takes place on the upper conveyor, curing in tunnel-like conveyors on the lower one [260]. However, flow production is not used exclusively for planar-type elements; it is also employed for stairs and linear components, for instance [265]. The different ways in which “series production” can be carried out these days are described in [252] and the factory production of four complete individual housing units every day using the latest techniques is described in [253]. Refs. [254] and [250] present more recent developments in


214

4 Factory production

the field of the production of elements for large-panel construction, and also discuss how battery moulds for walls can be integrated into flow production systems. Pretensioning forces must be resisted by the moulds themselves where prestressed elements are produced in the flow production method (see also section 4.4.2). Floor elements are particularly suitable for production on long lines [255]. For example, up until recently, precast planks for composite plank floors were produced almost exclusively on long lines, with compaction being carried out by vibratory carriages running below the bed or by external vibrators attached with quick-action couplings. However, the large factory floor areas required for this, the mobile workstations and the relatively long transport distances have seen a move towards pallet production in newer plants, with automatic stacking systems in the curing chambers (Fig. 4.1) [256].

Fig. 4.1

Flow production system for the production of precast concrete planks (Nuspl)

4.1 Production methods

215

Whereas conventionally reinforced hollow-core slabs are produced entirely on pallets by means of flow production, prestressed hollow-core slabs are produced almost exclusively in long lines [257]. We must distinguish between two fundamentally different manufacturing methods: x The slipformer functions like slipforming formwork and is pulled over the production line with a winch. The feeding units mounted on this operate with three filling and compacting stages (Fig. 4.2a). The lower machine unit can be changed to suit different cross-sectional forms as required. x The extruder works according to the recoil principle, so to speak (Fig. 4.2b). It braces itself against the ribbon of concrete it has produced and thus propels itself forward. In doing so, a very stiff concrete is processed which is pressed into the profiling zones by augers and at the same time is compacted by high-frequency vibration. This gives the concrete the necessary early strength required for this method of production and a high final strength. This production method represents the highest degree of mechanisation currently possible in precast concrete construction. The lines are cleaned by machines, the prestressing wires are laid automatically and the elements are cut to length with a fully automatic travelling concrete saw. This method was used, for example, to produce the 40 000 floor elements needed for the University of Riyadh (Fig. 4.3). Computer-assisted methods such as CAD (computer-aided design) for producing the working drawings and CAM (computer-aided manufacturing) have for some time now been employed in the production of such floor elements [92, 256, 258, 259, 261]. Considerable progress has therefore been possible in precast concrete construction recently in the three traditional areas of automation [262]: – Design and development (CAD) [263] – Production planning and control (PPC) with materials management [264] – Process sequencing (CAM) [265] and production data acquisition (PDA) [266] Fig. 4.6 illustrates schematically how the reinforcement can be fixed automatically by a computer-controlled robot. Double-T floor units (Fig. 4.5), T-beams, I-beams and “inverted V-beams” (for sawtooth roofs) are produced in long lines, which are normally combined with prestressing beds. Developments here are in the direction of hydraulic or electromechanical adjustment of the mould [255]. Mould plotters, mould robots [267] and concrete distributors are already being controlled fully automatically via CAD. These days, the precast concrete components that are still produced on conventional table formwork are those unsuitable for batch production or those that need special moulds because of their size or prestressing requirements. Those components are mainly beams, prestressed double-T floor units, irregular wall panels and columns. Precast concrete components are not just ideal for production in series or batches. They can also be produced with complex geometries and surface finishes. Nevertheless, in order to achieve a consistent production process and minimise the work required for each

216


Fig. 4.2 Industrial production processes for hollow-core slabs [257]: (a) slipforming (b) extrusion (c) mandrels

217

4.1 Production methods

Fig. 4.3

Production of prestressed hollow-core slabs by means of extrusion

individual component, CAM methods have already been used on many occasions for milling the surface geometry. To do this, “permanent formwork” – generally made from polystyrene – is produced by the computer-controlled milling machine and placed as a lining in the mould. Almost any surface geometries can be modelled in this way. The side panels to the mould must be correspondingly deeper in order to accommodate the lining. When using polystyrene, however, a slightly rough surface finish must be expected because individual beads of polystyrene are lost from the surface during the milling work.

Fig. 4.4 slabs

Loading of prestressed hollow-core

218


Fig. 4.5

Production of double-T elements in long lines (Olmet)

Fig. 4.6

Fixing the reinforcement for a floor unit with a computer-controlled robot [256]

Fig. 4.7 shows the mould and a finished element for an HGV test track. The random surface was calculated with a CAD system and transferred to the milling machine for the permanent formwork. Every element in this test track is unique.

4.2 Types of concrete in precast concrete construction

Fig. 4.7

4.2

219

Precast carriageway for an HGV test track with milled mould lining (Zu¨blin)

Types of concrete in precast concrete construction

We have seen major developments in concrete technology in recent years. The following types of concrete are acceptable to the building authorities and can be used without restrictions: – Normal-strength concretes up to grade C50/60 [270] – High-strength concretes up to grade C80/95 [292] – Lightweight concretes up to grade LC60/66 – Self-compacting concrete according to the 2003 DAfStb directive on the subject [297] The following types of concrete may only be used in conjunction with a National Technical Approval or Individual Approval: – High-strength concretes of grades C90/105 and C100/115 – Steel fibre-reinforced concretes [293] A DAfStb directive is available for this latter type of concrete, but in draft form only at present. The latest developments are taking place in the following fields: – Ultra-high-performance concretes (UHPC) – Textile-reinforced concretes (TRC) [291] (see also section 2.4.5). In addition, there are very many special concretes such as: – Impermeable concrete – Acid-resistant concrete – Coloured concrete – Concrete with high frost resistance – Concretes reinforced with glass, synthetic and other fibres [298] – Combinations of the above types of concrete [295, 296] The types of concrete interesting for precast concrete construction will be briefly described below. Of course, concrete technology now covers a wide field, and the interested reader should refer to the publications mentioned above.

220


All the aforementioned types of concrete are perfect for use in precasting plants. This is because the concrete is mixed and used directly in such plants, the moulds are already available, the concrete is easy to work and the working conditions are ideal. Therefore, almost all the first uses of such concretes on a commercial scale have taken and continue to take place in precasting plants or with precast concrete elements. This trend offers the precast concrete construction industry good opportunities now and in the future. 4.2.1

Processing properties

Concrete for the production of precast concrete elements often has to satisfy different requirements to those of in situ concrete. In the precasting plant, properties important on a building site, e.g. long working time or slow heat development, are irrelevant, indeed even undesirable. First of all, the wet concrete should be easy to pour, should not remain in the skip or stick to the chute. It should then not segregate in the mould before it has hardened, which is something that constituents of the mix with different densities tend to do: light aggregates float, heavy aggregates settle, foamed concrete (see section 4.2.4) at the base of wall panels concreted vertically becomes denser and heavier than that at the top. And water, too, the lightest constituent in normal-weight concrete, should not be allowed to separate out and lead to bleeding. These requirements are fulfilled by a concrete mix that stiffens rapidly and does not leave the constituents enough time to segregate, e.g. by choosing a rapid-hardening cement that retains the water, by limiting the aggregate size to H16 mm, or by using concrete with a low water content. Rapid hardening is also important when the element has to undergo heat treatment after concreting (see section 4.3.1) because the interim storage time is prolonged by late setting. The short time from mixing the concrete to pouring it into the mould and the additional compaction options offered by factory production enable the use of a stiff to plastic concrete consistency. Less mixing water is therefore required, which results in many advantages (see Table 4.1). 4.2.2

Strength

Early demoulding so that the moulds can be used again without delay requires a concrete that sets quickly. Once a compressive strength of about 5 N/mm2 has been reached, the skin of cement laitance on the surface is no longer pulled off during demoulding. But higher strengths are usually necessary for lifting the elements out of their moulds and storing them until curing. For example, transport fixings require a certain concrete strength. The safe working loads for such fixings are specified for a compressive strength of 15 N/mm2 (see also DIN 1045- 4). If such a strength cannot be guaranteed, then the safe working loads must be reduced accordingly, or the anchor embedment depth increased.


Table 4.1

221

How the mixing water effects the concrete

Less mixing water results in...

Advantages

rapid setting

Early trowelling/floating of upper concrete surface possible, better condition for heat treatment of concrete

stability of fresh concrete

Early removal of mould side panels possible

early strength

Early demoulding and early curing possible

fewer pores

Dense concrete, solid concrete

less shrinkage

Dimensional accuracy, no cracks

Higher compressive strengths are required for components that must be prestressed (see section 4.4.2). DIN 1045-1 prescribes the permissible bond stresses for anchorages depending on the compressive strength of the concrete at the time of prestressing. Particularly lightweight concretes such as foamed concrete (see section 4.2.4) represent an exception here because they exhibit much lower strengths than other types of concrete. Their compressive strength depends primarily on their density and for the types used for reinforced concrete, i.e. density i 1.5 kg/dm3, is about 3 N/mm2 after one day, 9 N/mm2 after seven days [268]. The early strength requirement derived from the needs of production is so high that the final strength necessary for structural and constructional reasons is in most cases guaranteed. Concrete grade C35/45 or C45/55 is used as a rule. It is therefore mostly the strength needed at the time of demoulding that governs the concrete mix, not the final strength. Manufacturers therefore use rapid-hardening cements (42.5 R or 52.5 R), reduce the water-cement ratio by using a high cement content (350 kg/m3 and more) or a plasticiser, work with a stiff consistency (sometimes so-called earth-moist), or – but rarely – add an accelerator. The plasticisers available today enable the production of concretes with very low watercement ratios (w/c 0.25– 0.35), which means that concrete strengths up to grade C70/85 are readily possible. These high-strength concretes had already been in use in precast concrete construction for a number of years – albeit not taken into account in structural analyses – before they were included in DIN 1045-1. Silica fume must be added (preferably in the form of a suspension) to achieve strengths up to grade C80/95. The silica fume brings about a further increase in strength of approx. 20 %, but no increase in the elastic modulus, and more early shrinkage. The latter can cause shrinkage cracks in the new concrete component. Careful curing is therefore vital. Curing times for high-strength concretes should be approx. 1–2 days longer than normal-strength concretes. Water loss must be prevented, indeed, water may even need to be added during curing. The mechanical properties of high-strength concretes cannot be deduced from the properties of normal-strength concretes by linear extrapolation because high-strength concretes are far more brittle. Fibres, mostly steel fibres, can be added to high-strength concrete in order to improve its deformation capacity. Besides a lower ductility, high-strength con-

222


cretes also exhibit a lower fire resistance, or rather in the event of a fire the concrete cover spalls off earlier [301]. The DAfStb directive on high-strength concrete calls for mesh reinforcement within the concrete cover in order to prevent losing the entire concrete cover and exposing the longitudinal reinforcing bars. Polypropylene fibres (approx. 2–3 kg per 1 m3 concrete) are steadily replacing the mesh reinforcement. These fibres melt as the temperature rises and the ensuing voids reduce the water vapour pressure which would otherwise lead to the undesirable, abrupt loss of the concrete cover. Very high cement contents were sometimes used in the early years of these developments, which led to a dough-like consistency and the need for intensive mixing; the sensitivities of the concrete, excessively early setting, etc. were relatively high. Replacing part of the cement by pulverised fuel ash (PFA) has a good effect on consistency and workability. These days, the workability of high-strength concrete is very similar to that of normal-strength concrete, but intensive compaction is especially important, and the intensity of compaction must be increased. Currently, high-strength concrete is mainly used for components in compression [106, 299, 300]. Ref. [292] contains an extensive bibliography on this subject. Reducing the w/c ratio, or the water/binder ratio, further has resulted in the development of ultra-high-strength concrete. The cement content is approx. 600 –1000 kg/m3, with microsilica contents of 250 kg/m3. The binder content is approx. 500 kg/m3 and is roughly double that of normal-strength concrete. Ultra-high-strength concretes are generally produced with an aggregate size of max. 2 mm. Heat treatment increases the strength of the concrete even further. For example, strengths of up to 200 N/mm2 can be achieved by curing at temperatures of up to 90 hC, and up to 800 N/mm2 with temperatures of up to 400 hC. All the details given above regarding the properties of high-strength concrete apply to ultra-high-strength concrete even more so. Ultra-high-strength concrete is frequently referred to as ultra-high-performance concrete (UHPC) because it is extremely durable as well as being extremely strong [302]. Its very dense microstructure results in extremely good resistance and in that respect UHPC is particularly useful for drainage installations with high acid contents. In addition, an attempt is being made to exploit the high compressive strengths for prestressed components [303] as well as compression members [304]. The aim of some developments is to minimise the steel reinforcement in the concrete, indeed, even eliminate it completely! The first applications, each completed with an Individual Approval, have already been built, e.g. the Ga¨rtnerplatz Bridge in Kassel, Germany. Fig. 4.8 shows the ultra-high-strength concrete being poured into the mould to produce the bridge deck. Ultra-high-strength concrete has already been used in France and Canada, where it is marketed by Lafarge under the trade name DUCTAL. The first footbridges in Canada and Japan have now been joined by unreinforced façade panels (see also section 2.4.5).


Fig. 4.8

4.2.3

223

Bridge deck made from ultra-high-strength concrete: (a) production, (b) erection (ELO)

Self-compacting concrete (SCC)

In contrast to the ultra-high-strength concretes, a DAfStb directive on self-compacting concrete is already available, which means that this type of concrete is already being widely used [297]. A commentary on the directive, background information and practical advice can be found in [305–307]. The flowability of this concrete is achieved by adding a very efficient plasticiser and the self-compacting effect through a suitable binder/aggregate ratio and a special grading curve. Compared to normal-weight concrete, self-compacting concrete is characterised by the fact that in the wet state it flows under the action of gravity until it finds an even level and in doing so releases all the entrapped air. Care must be taken during concreting to ensure that the SCC can flow over a certain distance in order to lose all the air. The strengths achieved are as for normal-weight concrete. However, as with all special concretes, careful curing is important. A sealed, level mould is vital. Construction joints must be provided in components with changes of level. The second concreting operation can be carried out after just 1–2 hours. Sloping surfaces constitute a problem. Nevertheless, self-compacting concrete has a number of important advantages, especially for precasting plants: – No compacting necessary – Less noise in the plant – Extremely good fair-face finishes – Very good encasing of cast-in parts – Heavy reinforcement possible – Better mould accuracy due to the elimination of vibration One example of the application of SCC is the production of the track for the Transrapid 2010 maglev railway system by Zu¨blin. The high accuracy requirements led to the decision to use self-compacting concrete for producing the slabs. A surface flatness of e0.5 mm over an area measuring 2.80 x 6.12 m was achieved. Fig. 4.9 shows the concrete flowing during the concreting operation and a finished surface.

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Fig. 4.9 Bridge deck made from self-compacting concrete: (a) concreting, (b) finished surface of deck (Zu¨blin)

Self-compacting concrete in the form of high-strength and lightweight concretes is currently the subject of research [295, 296]. Here again, the development is in the direction of prestressed components so that the high strength can be fully exploited. 4.2.4

Fibre-reinforced concrete

Attempts to eliminate conventional steel bar reinforcement by adding fibres (wood, glass, steel or synthetic, in the past also asbestos) to the wet concrete are not new. This is, however, only successful when the tensile bending strength of the concrete in the uncracked state is sufficient to resist the tensile stresses that occur, e.g. in lightweight roof tiles or small vessels, in pipes (steel fibres), or in façade facing leaves (AR glass fibres). In this type of concrete the fibres bridge over flaws in the concrete such as shrinkage cracks, but they cannot replace the loadbearing reinforcement of reinforced concrete. However, there has been significant further development in steel fibre-reinforced concrete. It seems that the steel fibres serve principally to improve the so-called post-failure behaviour, but can also replace part of the tensile bending reinforcement. In the meantime, steel fibres have become an essential ingredient for improving the ductility of the newly developed high-strength and ultra-high-strength concretes. Furthermore, there have been some individual attempts to replace the conventional steel bar reinforcement by steel fibres. Currently, it is only possible to replace the shear links by applying a prestress [308, 309]. A National Technical Approval for such an application has already been issued. If the DAfStb directive on steel fibre-reinforced concrete, currently in draft form, is soon accepted by the building authorities, then further significant developments in the field of steel fibre-reinforced concrete will certainly follow. Glass fibre-reinforced concrete has been widely used in the UK and USA in particular [281]. It can be used for producing self-supporting façade elements with approx. 15 mm thick walls using a manual sprayed concrete technique. The elements can only support their own weight and transfer wind loads directly to the loadbearing structure,


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Fig. 4.10 Lightweight noise barrier made from textile-reinforced concrete (Zu¨blin) (a) Textile-reinforced sandwich element; (b) Finished noise barrier

i.e. they are not primary loadbearing elements and so must be used in conjunction with structural steelwork or in situ concrete. Their low self-weight and the diverse architectural design options mean that they can be used for refurbishment work and for adding extra storeys to existing buildings. All the surface finishes possible with conventionally reinforced precast concrete components can be used. The method of production is, however, similar to that of a synthetic material (GFRP). Synthetic fibres are also becoming important as well as steel fibres – and in Germany, too. One main problem with their use is their lack of durability in the concrete in many instances. However, this does not apply to polypropylene fibres, which are highly alkali-resistant. As mentioned above, these fibres are therefore used for improving the behaviour of high-strength concretes in fire. Another significant development is the use of textilereinforced concrete, which is explained in more detail in section 2.4.5. The textile reinforcement is based on developments with short glass fibres and makes use of alkali-resistant types of glass as well as carbon or synthetic materials (polypropylene) for producing yarns which are then made into woven fabrics or braiding. We have already witnessed the first applications, carried out with Individual Approvals. One example is the development of a lightweight noise barrier by Zu¨blin. This is a sandwich element measuring 66 q 530 cm and consisting of 10 –15 mm thick textile-reinforced concrete facings and a mineral wool core (Fig. 4.10). The reader is also referred to the DBV leaflet on glass fibre-reinforced concrete (in German only). 4.2.5

Coloured and structured concrete surfaces

Many diverse architectural design options are available for the surfaces of precast concrete elements. Apart from the colour, is also possible to change the structure of the surface, which, however, must be taken into account when designing the concrete mix. The reader is referred to the FDB leaflet on fair-face concrete surfaces for information regarding tenders for and assessment of such surfaces.

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If the concrete surface is not treated or worked in any way after demoulding, then it is the outermost layer of the concrete, the cement laitance, that is solely responsible for the appearance of the concrete. The properties of the coarse aggregate are then irrelevant and only the constituents of the cement paste – sand, cement and water – need to be chosen to suit the requirements. A high water-cement ratio produces a light-coloured surface. The colour depends on the type of cement (blastfurnace cement w light grey, white cement w off white, oil shale cement w brownish) and the colour of the sand. It can also be varied by adding pigments. Synthetic inorganic pigments, especially iron oxide pigments in the three primary colours red, black and yellow, are mostly used for colouring concrete [269]. Brown colours can be produced by mixing the primary colours. Chromium oxide green is used for producing shades of green, titanium dioxide for shades of white. Cobalt blue is available as a blue pigment, which, however, like chromium oxide, is very expensive. The pigments are available in powder, liquid (i.e. slurry) or granulated form [284]. Organic pigments, like those used for producing lacquers, are useless for concrete because of their poor light-fastness and because they would be decomposed by the alkaline concrete. Carbon black, which is frequently used for producing very dark concrete bricks and blocks, fades in sunlight and so a permanent colouring cannot be guaranteed. However, the mineral pigments iron oxide (for shades of yellow, red, brown and black) and chromium oxide (for blues and greens) are light-fast and alkali-resistant. The use of white cement results in more vivid colours with a lower pigment content, e.g. only a tenth of that needed with grey cement. It should be remembered that the subsequent lime secretions on the surface of the concrete (efflorescence), which are normally washed away by the rain after a few years, are more noticeable on darker surfaces. This effect is caused by the migration of the calcium hydroxide (Ca(OH)2) in the pore water from the interior of the concrete to the surface. There, the deposits combine rapidly with the carbon dioxide (CO2) in the air to form limestone (calcium carbonate, CaCO3) which is not readily soluble in water and therefore creates light-coloured patches or streaks on the concrete [270]. The application of a water-repellent coating can reduce these blemishes (see section 4.3.3). Structured surfaces or segmented components are the best solutions for disguising irregularities in the surface roughness and colour of the concrete surface or efflorescence (see section 2.4.2). Research findings on the subject of efflorescence are summarised in [285]. The colour of the cement paste also determines the appearance of concrete components with finely brushed and washed or sandblasted surfaces (see section 4.3.2). However, some of the coarse aggregate, the gravel or chippings, is exposed by this treatment and the result is a surface reminiscent to the fracture surfaces of sedimentary rocks. A concrete mix with a constant grading curve [271], and in some circumstances selected coarse aggregates, is best for such a finish. Where the aggregate is exposed to a greater depth, the typical exposed aggregate look, then it is primarily the coarse aggregate that determines the appearance. Gravel or chippings with a certain colouring and size are then necessary and should account for 50 – 60 % of the aggregate with gap grading (e.g. 2–8 mm). The colour of the exposed gravel or chippings can be emphasized by using white cement or by adding a pigment to the bin-

4.3 Producing the concrete in the factory

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der. White cement should be used with a whitish aggregate, appropriate pigments added to the cement matrix for coloured aggregates. Efflorescence is less visible on exposed aggregate surfaces with a coloured aggregate than is the case with finely brushed and washed or unworked concrete surfaces. Types of concrete with expensive coloured aggregates can be used purely as a thin facing layer in order to reduce the cost of materials. Of course, this solution can only be used in conjunction with horizontal precasting. A concrete distributor with an appropriately fine distribution will be needed. The coarse aggregate must be firmly anchored in the cement matrix if the hardened concrete is to be worked with stonemason’s tools afterwards (see section 4.3.2). A strong concrete with a constant grading curve is best for such finishes. Concrete for profiled moulds (e.g. ribs or textures) is generally the same as that used for the normal loadbearing concrete. Only in the case of narrow profiles is it necessary to reduce the maximum size of the aggregate accordingly. 4.3

Producing the concrete in the factory

4.3.1

Heat treatment and curing

The hardening phase of the precast concrete components depends on how much time is allocated to the concreting to demoulding phase in the production plan. If this is very short, e.g. 4 hours, then the hardening is best guaranteed by the application of heat. In this situation a type of concrete with a long setting time, so that it remains workable, is often used and then afterwards heat is applied as necessary to speed up the reaction of the cement in the concrete until the desired strength is achieved. The DBV status report on heat treatment and the DAfStb directive on this subject describe the methods available for doing this (see also [322]). The simplest way is to use steam (actually hot water vapour), which apart from a boiler needs no other large equipment – essentially only tarpaulins or other forms of covering. Care should be taken to ensure that the surface of the concrete is not washed away by dripping condensation and that the temperature below the tarpaulins etc. is identical everywhere. Several steam lines at different places will be needed for long components. Hot-air treatment works in a similar way to the steam method. As with all the other heat treatment methods described below, it is important to make sure that the surface of the concrete does not dry out, which is achieved by covering with sheeting or spraying with water. Heating with infrared lamps is carried out in heating chambers and the advantage of this method is that the infrared radiation only heats up the object concerned, less energy is lost to the surroundings. In addition, thermostats allow the system to be easily regulated according to the temperature of the concrete [310]. Combinations of various methods are particularly helpful with large components. The mould is heated by a heat transfer medium (oil, steam, water) or electric heating wires and the top side is insulated to retain the heat.

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Hardening the concrete at a high temperature (e.g. above 30 hC) increases the amount of the reaction products of the cement, which boost the early strength, but reduces the number of cement bonds contributing to the final strength. Heat-treated concrete components therefore exhibit a lower final strength than components made from an identical concrete mix but stored in cool conditions. This effect is enhanced if the components are not stored until they have set before starting the heat treatment [311]. Curve a in Fig. 4.11 shows the ideal progress of heat treatment [312]. According to this, a total storage time of about 10 hours is necessary. This is usually too long for practical operations. Indeed, the heat treatment is intended to speed up the work! A short period of heat treatment is therefore generally employed: Fig. 4.11 curve b and – with preheating of the wet concrete – curve c [315]. New studies have confirmed, however, that the durability of the concrete components can suffer in severe weather conditions [316]. Mind you, these studies were carried out after damage had been discovered on outdoor horizontal railway sleepers, which apart from the considerable dynamic loads were also exposed to moisture and frost effects that would never affect vertical façade elements or interior components. Curing the concrete also includes the cooling phase. Moisture treatments improve the density of the concrete surface and hence its resistance to penetration by carbon dioxide and pollutants, also water, which at the same time increases the resistance to frost and abrasion. Curing must be carried out according to the corresponding DAfStb directive. However, it is rare for a moisture treatment to be completed because delivery deadlines, lack of storage space or “overbooked” lifting equipment make this impossible. A curing membrane is sometimes sprayed onto the wet or fresh concrete. Before using this method, however, it is important to clarify whether paint is to be applied to the concrete surface at a later date because the membrane impairs the bond of water-based dispersion paints. A suitable curing membrane will then be required, e.g. a finally dispersed acrylic dispersion, which is compatible with a solvent-based coating [317]. It is certainly the case that the curing of external components influences the durability of the components just as much as, for example, the concrete mix [318]. Clients who require more extensive curing measures than those described in the DAfStb directive – depending on the influences to which the finished component will be exposed – must define

Fig. 4.11 Time-temperature graph for heat treatment showing the various stages (according to [313, 314])

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these as separate items in the specification. The ongoing development of precast concrete component manufacture should therefore also consider the organisation and provision of the facilities required for curing such elements. 4.3.2

Working hardened concrete surfaces

Closely associated with curing are a number of methods of working the hardened concrete, first and foremost the newly cast concrete, which expose the aggregate. To do this, the cement laitance on the surface is removed by weak acids, sandblasting or jets of water (Fig. 4.12). The last of these requires the least effort. The cement laitance can be washed off of the surface of the concrete as long as it is still soft. This is one method of creating an exposed aggregate finish. We distinguish here between a light treatment that removes only a fine layer and emphasizes the colour of the cement paste, and more intensive treatment that exposes the coarse aggregate and allows this to dominate the appearance. Water-jetting is carried out as soon as the strength of the concrete allows by delaying the setting of the concrete at the surface to be treated. This is achieved by applying a surface retarder to the mould (negative method) or spraying it onto the concrete surface after demoulding (positive method). Rolls of paper precoated with a retarder can also be used. Carrying out this coating work in the factory guarantees a consistent coating thickness and hence a consistent depth of exposure. The paper is laid in the mould and removed prior to water-jetting. It is important to make sure that no folds or creases – due to saturation with retarder – form during concreting. The depth of the effect on the surface varies de-

Fig. 4.12 Working of concrete surfaces: (a) water-jetting a column, (b) sandblasting a window element, (c) flame cleaning for a spandrel panel element

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Fig. 4.13


Examples of surface finishes for precast concrete components

pending on the type of retarder and the coating thickness, enabling the cement laitance to be removed with a brush and jet of water once the concrete in the core of the component has hardened to such an extent that it can be removed from the mould. With some substances the concrete surface begins to harden as soon as the component is removed from the mould and comes in contact with the air, but others only after they are struck by the jet of water and the substance is diluted by the water, and with yet others the hydration of the cement is practically permanently inhibited. Sandblasting [319] of the previously hardened concrete surface achieves a similar appearance to light water-jetting but the surface of the exposed grains of aggregate are somewhat roughened by this technique and they lose their natural shine. This is of course not important with angular grains of coarse aggregate with naturally rough surfaces. Sandblasting requires adequate protective measures to protect the surroundings against dust, e.g. sandblasting tents. Quartz sand is still used for sandblasting concrete (in con-


231

trast to metals) because the abraded concrete itself contains the quartz sand substances that are harmful to the lungs. Washing the surface with acid has largely been replaced by surface retarder methods for health reasons [320]. The surface of the hardened concrete can be worked in other ways: flame cleaning [321] or mechanical stonemason methods such as bush-hammering, scabbling, grinding, etc. (see DIN 18500 “Cast stones”). What all these methods have in common is that they split the exposed grains of aggregate and allow these fractured surfaces with their strong natural colours to dominate the appearance. However, such techniques are very labour-intensive and are only used on special projects. Apart from grinding, they cause minor cracks in the remaining surface which afterwards should be sealed with a hydrophobic substance. Grinding can be used to create very high-class façades as good as natural stone. Large-format precast concrete façades require bulky and expensive grinding plant, however. Ref. [323] describes one new method of production for façade elements. A less expensive variation on the stonemason-like working of the surface is achieved by casting the concrete with ribs and subsequently breaking off the ribs with a suitable hammer. The fractured surfaces are coarser than with bush-hammering. Fig. 4.13 shows examples of finished surface structures. 4.3.3

Coating and cladding

Only in exceptional circumstances, e.g. severe chemical attack, is it necessary to coat concrete components to provide them with the necessary durability. Any coatings that are applied for architectural purposes should also improve the durability of the concrete, and the materials used must be alkali-resistant, light-fast (UV radiation) and water-resistant. And where the component is exposed to temperature fluctuations, then the coating must also be permeable for water vapour, i.e. repel water in liquid form, but allow water in vapour form to escape from the interior of the concrete. Siloxanes – an intermediate stage between the monomeric silanes and the silicones (the latter frequently used on concrete in the past) – satisfy this requirement, penetrate several millimetres into the concrete pores as a result of their small molecules and form a waterrepellent layer on the surface. This film is very thin and therefore invisible; the appearance of the concrete surface is therefore hardly changed. It is important to apply these coatings evenly and in sufficient quantities according to the manufacturer’s instructions so that the water-repellent effect is consistent over the entire surface and no patches form when the concrete is wet. They slowly degrade in UV light but this is hardly relevant in practice provided they have penetrated deep enough into the concrete (two coats are therefore usually recommended). We can therefore assume that a concrete surface coated in this way will need to be re-impregnated after about 10 years after removing all dust deposits and growths. Siloxanes degrade less on surfaces not exposed to direct sunlight. Acrylic resins offer more resistance to UV radiation than siloxanes. In addition, they prevent the infiltration of carbon dioxide from the air, which reduces the alkalinity of the concrete. It is primarily the alkalinity that protects the reinforcement against corrosion. Acrylic resin solutions can also be used in conjunction with siloxane impregnation. This results in a somewhat thicker coating that lends the surface a satin finish with a dar-

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ker and more vivid colouring, an effect that is particularly beneficial when using coloured concrete. Rain can wash off deposits of dust and dirt more easily from a surface treated in this way. Coating the surface with solutions or dispersions to which pigments have been added has an even stronger effect on the appearance of the concrete. Such coatings are known as sealants when applied as a very thin film (up to 0.3 mm). These contain just enough pigment to change the colour of the concrete but not conceal its texture, i.e. they are like a glaze finish. They can be used to compensate for fluctuations in the natural colour of the concrete surface and at the same time provide protection. The structure of the concrete surface remains visible. Opaque paints for concrete are about twice as thick and are available in a vast range of colours. They are mainly in the form of dispersions. The water vapour permeability becomes more important as the thickness of the coating increases, and the permeability of dispersions is greater than that of solutions with the same coating thickness. As mentioned in section 4.2.5, pigments for exterior paints should be alkali- and UV-resistant. These are generally pure mineral pigments, some of which are obtained from rare earths. Unlike when mixing different paints to obtain a certain shade, the colour from one batch to another cannot be kept identical when manufacturing colourants with pure pigments and so whenever possible only paints from the same batch should be used for a project in order to avoid discrepancies. Organic pigments, which are suitable for coating timber and metal, do not have the necessary alkali resistance. Coatings in bold colours are possible, but there is a risk that they will heat up to a greater extent than more muted colours and expose the components to more extreme temperatures, which in turn can lead to higher vapour pressures. Such bold colours should therefore be restricted to smaller areas only or be applied in the form of stripes. Far fewer demands are placed on coating materials for interior components because they do not need to be UV-resistant and in most cases can be renewed without the need for ex-

Fig. 4.14 Façade elements with ceramic tiles attached in the factory

4.4 Installing the reinforcement in the factory

233

tensive scaffolding. And indoors the water vapour permeability is less important than with external components. The coating materials for interior use are therefore correspondingly cheaper. Precast concrete components should not be painted until they have been installed in order to avoid soiling and damage during transport, storage and erection. Paints are not the only finishes possible; plaster, render and stone or ceramic tile finishes can also be completed in the precasting plant (Fig. 4.14). Such finishes always call for especially careful handling during transport and erection. If such finishes are not attached until after erection, they can be adapted to suit the actual as-built dimensions that result from the inaccuracies during construction, and also cover up minor damage. When attaching tiles or similar finishes to the finished precast concrete components, it is important to check that the adhesive is compatible with the release agent and will remain permanently elastic in order to prevent damage caused by the different thermal expansion behaviour of the materials either side of the adhesive joint. Panels of tiles should be kept suitably small, separated by permanently elastic joints. 4.4

Installing the reinforcement in the factory

4.4.1

Round bars and meshes

On average, about 20 % of the total cost of a precast concrete component can be attributed to the reinforcement. It should therefore be given due attention. In doing so, the requirements of the structural calculations on the one hand and the requirements of an economic reinforcement layout plus adequate concrete cover on the other must be taken into account. The requirements regarding the correct and adequate depiction of the reinforcement on the drawings for precast concrete structures are summarised in [324]. Fig. 4.15 shows the title block of a working drawing for a precast concrete component. It may be necessary, especially with corbels, notched beam ends or in the vicinity of loadbearing castin parts, to draw the reinforcement at a larger scale with parallel lines and all bends shown to scale (Fig. 4.16). Where bars cross or are placed directly alongside each other, it is essential to consider the fact that the actual outside diameter of a ribbed reinforcing bar is approx. 20 % greater than its nominal diameter. And where bars are placed in several layers (e.g. in corbels), it is also not only the final fit that is important, but also the “buildability”. FDB leaflet No. 5 (2005) (in German only) provides valuable information on the planning and drawing errors frequently encountered in practice. DIN 1045-1 defines the allowance for the concrete cover as well as the minimum dimension because of the potential tolerances. This enables a nominal dimension (w minimum dimension S allowance) to be defined. The allowance as well as the cover to the nearest concrete surface must therefore be specified on the drawings of the elements. Experience has shown that insufficient concrete cover is frequently caused by inadequate bar spacers. Only those bar spacers that comply with the DBV leaflet on bar spacers (see DBV “Concrete Best Practice”) should therefore be used. If with a rigid mould the bar spacers are too pliant (e.g. made from plastic) or are too heavily loaded and therefore squashed, then they

Element No. Cast-in partNo.

pcs./element Σ pcs. Designation

Cast-in parts Min. values for bending roller diameter dbr (reinforcing bar types III + IV) Hooks, bends, links, loops

Bent up, bent down, other bent forms

ds link cside

clink ds < 20 mm : dbr 1 = 4 ds ds ≥ 20 mm : dbr 1 = 7 ds

clink cside > 5 cm & > 3 ds : dbr 2 = 15 ds cside ≤ 5 cm or ≤ 3 ds : dbr 2 = 20 ds

Special dimensional tolerances (where different from DIN 18202 and DIN 18203) Surface finish Mould rough Trowelled Rubbed Floated

Chamfers, x = _____ cm Exposed aggregate, type ___ / colour ___ Worked finish, type ___ / colour ___ Rough for concrete topping

h g f e d c b a In- Date dex

Name

Revision

In- Date Client dex

Checked Production Erection _____

Concrete strength class

B____ ____ m3 LB____ ____ m3 B____ ____ m3 Density class ________ Compressive strength of concrete for transport ≥ ____ N/mm2

Special properties Water impermeability, ew ≤ ____ cm Air entrainer content, min./max. ___/___ % by vol. Other

Reinforcing steel

Prestressing steel

BSt____ ____ kg BSt____ ____ kg

St____ ____ kg St____ ____ kg

Concrete cover (nom. dim.) c= c= c=

Fire resistance rating

cm cm cm

F F

-A - AB

Follow advice in concrete cover leaflet!

Checking endorsement

Approval for production

The drawing, index ____, agrees with the checked drawing/structural calculations. Date

Signature

Date

Signature

Mould Reinforcement Concrete

Company

Client: Project: Site: Component:

Mould Reinforcement

No.

Part

Weight (t)

Scale 1: __, 1: __ Drawn Checked

Date

Signature

Structural No.

Project No.

Date

Drawing No.

Signature

Group leader Dept. Head

The drawing, the associated annexes, descriptions, calculations, etc. and their content are our intellectual property. They may not be reproduced, made available to unauthorised third parties or otherwise disclosed or used for purposes other than those for which they are intended without our permission. They must be returned upon request.

Sheet size

Fig. 4.15 drawing

Title block for a production


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Fig. 4.16 Reinforcing bars drawn with double lines and all bends to scale

return to their original shape after demoulding and push off the outer layer of cement laitance. And where there is a soft lining in the mould, which can be the case with textured finishes for façade panels, for instance, bar spacers can press into the lining. In such situations it may be necessary to suspend all the reinforcement from cross-beams. Precasting plants generally make use of reinforcing steel grade BSt 500 S in accordance with DIN 488. The fact that only weldable steel is used these days is a major advantage for precast concrete construction with its many cast-in parts for joints and connections. Reinforcing cages for linear elements such as beams and columns are generally assembled outside the mould, i.e. prefabricated, whereas the reinforcement for floor slabs is generally fixed directly inside the mould (double-T units, precast planks, hollow-core slabs). Assembling the reinforcement for beams or T-beams within the mould calls for open shear link cages to simplify the fixing of the longitudinal bars. So-called closer bars are then fixed on top to close off the open shear links. Shear links made up into cages with the help of welded longitudinal bars can be stacked more easily if the straight bars are placed outside the links (Fig. 4.17) [325, 326]. Fixing the reinforcing bars directly from a coil is a technique that is already being used in many precasting plants, especially for the smaller bar diameters (6–14 mm). There is no wastage and the work can be carried out practically without interruption. Several diameters are permanently available (Fig. 4.18). The processing costs per tonne of steel for the small bar diameters in particular are disproportionately high when compared with the larger diameters. The DIBt can issue approvals for reinforcing steel in coils [327], which means that there is nothing to stop the widespread use of this practice in Germany. The first publication by the manufacturers and the processing plants for reinforcing steel in coils shows just how hectic the developments are in this field [328]; the demand for reinforcing steel in coils increased more than 10 -fold between 1985 and 1990 [331] and has probably doubled

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Fig. 4.17 Stackable shear link cages: (a) stack of shear link cages with internal longitudinal bars, (b) stack of shear link cages with external longitudinal bars


Fig. 4.18 System for processing reinforcing bars directly from coils

again since then. Hot-rolled ribbed steel grade S 500 WR is approved for processing directly from the coil. The same application conditions apply as for grade BSt 500 S. Since 1986 there has also been an approval available for using “cold-worked stainless steel ribbed reinforcing bars of grade BSt 500 NR” directly from the coil for diameters up to 14 mm. This steel can also be welded to non-alloyed steels. Only the bond behaviour is relevant for the concrete cover; it is not necessary to increase this to take into account the environmental conditions. However, its high cost means that this type of steel is used in special cases only, e.g. filigree façade elements or starter bars in highly corrosive conditions. Processing reinforcing steel directly from the coil for helical column reinforcement has long since been standard for pipes, for example. This led to the development of similar machines for making up reinforcing cages for square or octagonal columns and rectangular beams, which, for example, was used successfully for reinforcing the precast concrete components for the University of Riyadh (Fig. 4.19). Besides the automatic straightening and cutting machinery with which the reinforcing bars are processed directly from the coil to form straight bars, there are also automatic link bending machines that turn the steel into finished shear links. The straightening and cutting machines used in precasting plants can frequently handle up to four different diameters simultaneously and continuously [329, 330]. Problems with the direct control of automatic bending plant are discussed in [332]. The trend seems to be towards the automatic process-controlled fixing of reinforcing bars. Such systems are already available for floor slabs, for instance (see Fig. 4.5). In the meantime, the first fully automatic welding stations fully integrated into the production line are appearing, too. These are used to weld bars (directly from the coil) al-


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Fig. 4.19 Reinforcing bars directly from the coil for shear link cages (Zu¨blin): (a) automatic welding machine with rectangular cage, (b) shear link cages for octagonal columns

ready straightened and cut to length into planar reinforcement in the form of “just-in-time bespoke meshes” [260]. 4.4.2

Prestressing beds

Precast concrete construction has made use of prestressed long-span floor slabs and roof beams for single-storey sheds in particular right from the early days of prestressed concrete construction. The well-known advantages of the latter, such as slender cross-sections, limited deflection, exploitation of the high tensile stresses possible in the steel and loading the concrete in compression, also enable the economic production of prestressed precast concrete [333–336]. In factory production, pretensioning in prestressing beds is practically the only method used (Fig. 4.20), and in Germany it is mostly seven-wire cold-drawn strands of strength class St 1570/1770 that are used for prestressing. The simplest approach is to lay straight wires in the long prestressing bed. Prestressed hollow-core slabs up to 150 m long can be produced in this way (see Fig. 4.3). Several long-span double-T units, e.g. for multi-storey car parks and the roof beams to singlestorey sheds, are produced mostly simultaneously in prestressing beds up to 80 m long. The optimum length depends on the daily output possible by the respective workforce, although in the case of prestressed hollow-core slabs it is the extruder that determines the output. Gravity foundations with anchorages must be provided as the prestressing bed abutments, which are braced against each other via the prestressing bed (Fig. 4.21). Modern precasting plants generally have prestressing beds with 3–5 MN prestressing force (approx.

238


135 kN/strand), but prestressing beds with up to 15 MN are necessary for bridge beams. Short prestressing beds for panels and double-T units are also used, especially for flow production systems, in which the prestressing force is braced against the stiff steel mould, which generally requires only minimal strengthening for this. The disadvantage of straight prestressing is that the prestressing force does not correspond to the bending moment diagram of a single-span beam. It is positioned too low at the ends of the beam and this leads to tensile stresses in the top here. One way of overcoming this problem is to place some of the strands in sleeves at the ends of the beam to prevent them bonding with the concrete. But if the strands are to follow the bending moment diagram, then bent (harped) strands are unavoidable. This is carried out either once in the middle, as was sensible with, for example, double-T units with long notches at the ends according to Fig. 4.22, or twice, e.g. at the quarter-points of the beam (Fig. 4.23). In duopitch roof beams the pitch of the top flange is used as the angle for harping the strands. In other precast concrete components, e.g. prestressed hollow-core slabs, the only way of solving the problem may be to include additional prestressing in the top. The hold-down anchors for bent strands are either anchored in the base of the mould or forced downwards hydraulically via cross-frames. Once the concrete has hardened, the hold-down anchors are released and the openings grouted. The elastic limit (fp0.1k) of the prestressing steel may not be exceeded at the harping points of the strands, and here the extreme fibre stresses for strands may be calculated with half the nominal diameter. This can lead to relatively large radii when exploiting the permissible prestressing bed stress (0.9 fp0.1k or 0.8 fpk). The jaws in the chuck assemblies should certainly have rounded edges and be made from a softer steel than the prestressing strands if at all

Fig. 4.20

Pretensioning in prestressing bed (schematic) [326]


Fig. 4.21

239

Prestressing bed abutments

possible. Trials with seven-wire strands have revealed that with typical harping point angles of up to approx. 10h, rounding diameters of 100 or 200 mm are possible (Fig. 4.24). The pretensioning is transferred to the concrete either by releasing the jacks positioned between the perforated plates of the anchorages and the abutments or by burning through the strands, although the steel loses more and more of it is strength as it heats up. In the latter method, the cutting process should be carried out so that the stresses are introduced as symmetrically as possible in both directions. When producing prestressed hollow-core slabs, the individual units are cut to length by sawing (see Fig. 4.3c). Eccentrically prestressed elements have an upward camber after demoulding, when only their self-weight is effective. Inaccuracies in the prestressing force and different elastic moduli at the time of prestressing lead to different offset dimensions, which are often dif-

Fig. 4.22

Prestressed double-T unit with bent (harped) prestressing strands

240

Fig. 4.23


Harping the prestressing wires

ficult to deal with during erection. The strength of the concrete at the time of transferring the pretension also has a major influence on creep and shrinkage. Prestressing bed elements are often heated to achieve the early strength necessary. The prestressing bed stresses do not alter when tensioning against the steel mould and heating the mould and wet concrete evenly. The situation is different in the prestressing bed with a zero-friction mounted mould. The heat treatment normally begins after the concrete has started to set. At this point in time the bond between the concrete and the prestressing steel is already beginning to take effect. When the concrete and the mould expand against the external abutments of the prestressing bed, the prestressing force is either wholly or partly transferred to the concrete cross-section at an early stage because the thermal expansion of the element relieves the relatively short unbonded length of the prestressing strand between end of element and anchorage. So the heat treatment may only begin after a period of storage, during which the bond stress takes effect. Otherwise, the elongation of the steel – and hence the prestressing bed stress – must be increased by an amount equal to the thermal expansion. This effect does not occur when there is a friction connection between precast concrete component and prestressing bed, as is the case with prestressed hollow-core slabs cast in long prestressing beds. It is important to take into account the higher relaxation losses of the prestressing steel caused by the higher temperature during the heat treatment of the concrete. These losses consist of, on the one hand, accelerated prestressing force losses due to relaxation and, on the other, the thermal loss of the initial prestress (Fig. 4.26). Relaxation values are given in the prestressing steel approval documentation. When introducing the pretension into an eccentrically prestressed beam, there is a tendency for the beam to curve upwards and support itself on its outermost bottom corners. With notched beam ends there is then the risk that the beam has already developed cracks in the corner of the notch by the time it is transferred to the storage yard (Fig. 4.25). This

Fig. 4.24 Prestressing strands, 7 No. H 4 mm wires, grade St 1570/1770, tensile strength values for various harping angles and various roller diameters


241

Fig. 4.25 Curvature after relieving an eccentric prestress

can only be avoided by including a soft intermediate pad in the mould beneath the support nib. Such notched beam ends with straight ends to the prestressing strands in the bottom must be reinforced like conventionally reinforced notched beam ends (see section 2.6.2). The main structural feature of pretensioned prestressed precast concrete components is the direct transfer of the prestress at the ends of the component. This is achieved via the bond between the prestressing strand and the surrounding concrete and is amplified by the so-called Hoyer effect in which the cross-section of the strand widens because the prestress decreases towards the end of the component and thus increases the lateral compression on the prestressing strand [147]. The various bonding properties of the prestressing wires must be considered here. The pretension must be applied at a very early stage in order to avoid delaying the demoulding of the precast concrete component. The permissible bond stresses based on the concrete strength according to DIN 1045-1 Table 7 must always be taken into account in the analysis of the prestressing force transfer. If no cracks occur in the concrete cross-section within the calculated transfer length for the prestressing force, then further analyses are unnecessary. This is normally the case. If, however, only a minimal prestress is desired and the amount of conventional reinforcement is increased, then coping with the tensile force by way of reduced bond stresses must be checked [338–340, 346]. The results of the first studies into the use of self-compacting concrete in conjunction with prestressed components are now available; the studies are ongoing [341]. The tests did not reveal any significant differences between self-compacting and normal-weight

Fig. 4.26 Loss of prestress due to accelerated relaxation during heat treatment

242


Fig. 4.27 Tensile stresses in prestress transfer zone

concrete, which means that in the future we can expect to see self-compacting concrete being used in prestressed concrete construction as well. Tensile forces, as well as bond stresses, occur in the concrete due to the spread of the load in the prestressing force transfer zone. We must distinguish between bursting, tensile splitting and end face tension effects [337]. In roof beams and double-T floor units, the resulting tensile stresses are handled by shear links. In prestressed hollow-core slabs the special production process precludes the use of shear links. In this case adequate concrete cover in conjunction with a high concrete strength, which is achieved with extrusion, enable the tensile strength of the concrete to be exploited to resist these tensile forces with an adequate factor of safety. For this reason, elements without conventional reinforcement which are pretensioned in the prestressing bed require calculations for bending failure and flexure-shear failure plus an analysis of shear-tension failure, i.e. a limit to the inclined primary tensile stresses in the zone without bending cracks and an analysis of the anchorage capacity. Many tests on prestressed precast concrete floor slabs have revealed that the shear-tension capacity is always more significant than the flexure-shear failure in such elements [348]. The National Technical Approvals for prestressed precast concrete floor slabs include the appropriate analyses. 4.5

Quality control

Satisfying the requirements placed on the product “precast concrete component” is the topmost priority of any precast concrete manufacturer. Defining the quality of a product right from the very start is gradually becoming an intrinsic element in a quality management system based on the DIN EN ISO 9000 set of standards, which these days is normal for the management systems of manufacturers. The quality management system controls all the steps in the processes surrounding a precast concrete component, from receiving the order to providing after-sales services for components already delivered and in use. The quality control measures performed on the precast concrete component itself are only part of a set of diverse activities. The measures ensure, in the form of conformity checks, that the requirements placed on the manufacturer, the method of production and the con-

4.5 Quality control

243

crete produced are in compliance with the various regulations (DIN EN 206-1, DIN 1045 parts 1– 4, DIN 1048-5, DIN EN 12350 parts 1–7, DIN EN 12390 parts 1– 4, etc.). Conformity checks take place at the levels of production control and monitoring. Production control looks at internal procedures. The manufacturer’s own quality control is the prime focus here, which is recorded, e.g. in the factory production control manual, according to defined tests and inspections. Different requirements are placed on the monitoring and certification of precast concrete components depending on the intended usage of the component and the requirements placed on it by the building authorities (see Construction Products List A, B or C). In Germany the monitoring takes place in accordance with the quality assurance associations of the federal states, which are approved by the supreme building authority. The results of the monitoring visits every six months are recorded in test certificates and monitoring reports. According to the building authority status of the precast concrete components, this “external monitoring” is expressed in the following ways: – product certificate (P) (no building authority requirements), – conformity certificate () (for applications covered by building authority requirements – see Construction Products List A), or – certificate covering factory production control (2S) (prescribed by European Construction Products Directive). Quality control is carried out according to DIN 1045, which distinguishes between concrete categories 1 and 2. Category 1 covers concretes for secondary purposes, the monitoring of which is carried out by the manufacturer only. Concretes belonging to category 2 may only be produced under the control of a concrete specialist, e.g. “E certificate”. The production control is carried out by the manufacturer and covers the following areas: – Selection of raw materials – Concrete mix design – Production of concrete using materials subjected to QA measures (e.g. to DIN EN 206-1), – Factory production control (DIN 1045- 4) – Checks on the finished product The production control is monitored and certified at regular intervals by the aforementioned monitoring bodies. A German conformity mark () on the delivery document for a precast concrete component indicates to the recipient that the precasting plant supplying that component is subjected to external monitoring and complies with requirements placed on its manufacturing operations.

5

References

[1] Gaede, K.: Fertigteile aus Beton und Stahlbeton. Beton-Kalender 1958, Teil II, pp. 271–291. [2] Beck, H., Schack, R.: Bauen mit Beton- und Stahlbetonfertigteilen. Beton-Kalender 1972, Teil II, pp. 159–256. [3] Paschen, H.: Das Bauen mit Beton-, Stahlbeton- und Spannbetonfertigteilen. Beton-Kalender 75/82, Teil II, pp. 575–745 75/82, Teil II, pp. 533–696. [4] Koncz, T.: Handbuch der Fertigteilbauweise. Bauverlag, Wiesbaden, 1973, vol. 1/73, vol. 2/67, vol. 3/70. [5] Fachvereinigung Deutscher Betonfertigteilbau: Betonfertigteile fu¨r den Wohnungsbau. Verlag Bau S Technik, Du¨sseldorf, 2002. [6] Fachvereinigung Deutscher Betonfertigteilbau: Betonfertigteile im Geschoss- und Hallenbau (Verlag Bau S Technik GmbH, Du¨sseldorf, new edition 2009). [7] Fachvereinigung Deutscher Betonfertigteilbau: Fassaden – Architektur und Konstruktion mit Betonfertigteilen. Verlag Bau S Technik, Du¨sseldorf, 2000. [8] Fachvereinigung Deutscher Betonfertigteilbau: Ausbaudetails – Entwurfshilfen fu¨r den Fertigteilbau. Verlag Bau S Technik, Du¨sseldorf, 2002. [9] Beton S Fertigteil-Jahrbuch. Bauverlag, Wiesbaden, published annually. [10] Hahn, V.: Systembau aus Stahlbetonfertigteilen und Zusammenarbeit mit dem Architekten. Presentation, Betontag 1973. [11] Hahn, V.: Hat Industrialisierung im Bauwesen noch eine Chance? Der Architekt 1983, No. 10. [12] Berufsfo¨rderwerk fu¨r die Beton- und Fertigteilhersteller. Handbuch: Betonfertigteile, Betonwerkstein, Terrazzo. Beton-Verlag, Du¨sseldorf, 1991. [13] Bindseil, P.: Stahlbetonfertigteile. Werner-Verlag, Du¨sseldorf, 1991. [14] Bruggeling A. S. G., Huyghe G. F.: Prefabrication with Concrete. Balkema, Rotterdam, 1991. [15] Kotulla, Urlau-Clever: Industrielles Bauen – Fertigteile. expert verlag, 1987. [16] Cziesielski et al.: Lehrbuch der Hochbaukonstruktionen. Teubner, Stuttgart, 1991. [17] Meyer-Bohe, W.: Geschichte der Vorfertigung. Zentralblatt fu¨r Industriebau 1972, No. 5, pp. 186–191. [18] Ku¨hn, G., Go¨ring, A., Beitzel, H.: Neue Technologien fu¨r die Baustellen der Zukunft, Band I: Hochbau. Federal Ministry for Regional & Urban Planning, 1976, No. 04.018. [19] Rausch, H.: 10. Deutscher Fertigbautag – Ru¨ckblick und Ausblick. BMT Fertigbau 1985, No. 11, pp. 420-426. [20] Junghanns K.: Das Haus fu¨r alle. Zur Geschichte der Vorfertigung in Deutschland. Ernst & Sohn, Berlin 1994. [21] Breitschaft, G.: Harmonisierung technischer Regeln fu¨r das Bauwesen in Europa. Beton-Kalender 1994, Teil II, pp. 1–17. [22] DIN: Bauen in Europa, Beton, Stahlbeton, Spannbeton. Beuth Verlag, Berlin 1992. [23] Litzner, H.-U.: Grundlagen der Bemessung nach EC2, Vergleich mit DIN 1045 und DIN 4227. Beton-Kalender 1994, Teil I, pp. 671–864. [24] Kordina, K. et al.: Bemessungshilfsmittel zu EC 2 – Teil 1, Planung von Stahlbeton- und Spannbetontragwerken. DAfStb No. 425, 1992, 2nd ed.. [25] DBV: Beispiele zur Bemessung nach Eurocode 2. Bauverlag, Wiesbaden, 1994. [26] Meyer, H.-G.: Europaïsche Normen fu¨r Beton-Herstellung und Verarbeitung. Betonwerk S FertigteilTechnik 1993, No. 8, pp. 67–72. [27] Schießl, P.: Europaïsche Normen fu¨r Betonstahl und Spannstahl und europaïsche Regelungen fu¨r Spannverfahren. Betonwerk S Fertigteil-Technik 1993, No. 9, pp. 64–72.


246

5 References

[28] Mo¨nnig, F.: Anwendungsregeln und Normen fu¨r Hohlplatten in Deutschland und Europa. Vortrag, Stuttgart, 1993. [29] Leitfaden fu¨r die CE Kennzeichnung von konstruktiven Fertigteilen. Fachvereinigung Deutscher Betonfertigteilbau e.V., 2007. [30] Springborn, M.: Inverkehrbringen und Verwendung von Bauprodukten – die Bauproduktenrichtlinie und ihre Umsetzung. DIBt-Mitteilungen 1/2007. [31] Schwerm, D.: Die europaïschen Produktnormen fu¨r das Bauen mit Fertigteilen – praktische Umsetzung. Beton Fertigteil-Jahrbuch 2008. [32] Furche, J.: Neue Regelungen fu¨r den Fertigteilbau – Produktnormen fu¨r Elementdecken und andere Bauteile mit Gittertra¨gern, 52nd Betontage 2008, pp. 98–99. [33] Stengler, W.: Planungs- und Konstruktionsgrundsa¨tze. Betonwerk S Fertigteil-Technik, Fertigteilforum 1979, No. 2–4 & 8. [34] Polonyi, S.: Konstruktionsirrtu¨mer. Bauwelt 1978, No. 23, pp. 869–873. [35] Kerschkamp, F., Portmann, D.: Allgemeine Grundsa¨tze zur Maßkoordinierung im Bauwesen. DIN 18000 report, NA Bau 1988. [36] Toleranzen im Hochbau nach DIN 18202, pub. by ZDB, 2007. [37] Ertl. R.: Toleranzen im Hochbau, 2nd, updated & revised edition. Rudolf Mu¨ller Verlag, Cologne, 2008. [38] FDB leaflet No. 6: Passungsberechnungen und Toleranzen von Einbauteilen und Verbindungsmitteln, 2006. [39] Tiltmann, K. O.: Was kostet die Genauigkeit im Betonfertigteilbau? Baugewerbe 1977, No. 5, pp. 21–26. [40] Tiltmann, K. O.: Toleranzen bei Stahlbetonfertigteilen. Mu¨ller Verl.-Ges., 1977. [41] Paschen, H., Sack, W.-M.: Maßtoleranzen und Passungsberechnung im Stahlbetonskelett-Fertigteilbau. Bauverlag, Wiesbaden, 1980. [42] Paschen, H.: Bewertung und Behandlung von Maßtoleranzen im Fertigteilbau. Betonwerk S FertigteilTechnik 1981, No. 10, pp. 1–13. [43] Ostheimer, H.: Einfu¨hrung in das Genehmigungs- und Erlaubnisverfahren im Transportbereich. Kaisser Verlag, Salach, 1981. [44] Hahn, H., Sack, M., Steinle, A.: ZBLIN-HAUS. Karl Kra¨mer Verlag, Stuttgart, 1985. [45] Kordina, K., Meyer-Ottens, C: Beton Brandschutz Handbuch, 2nd ed. Verlag Bau S Technik, 1999. [46] Seiler, H.: Brandschutz im Industriebau. Betonwerk S Fertigteil-Technik 1983. [47] Kordina, K., Richter, E.: Zur brandschutztechnischen Bemessung vorgespannter Fertigteile. Betonwerk S Fertigteil-Technik 1984, No. 8, pp. 540–546. [48] Hasenkru¨ger, E.: Bauen mit Betonfertigteilen – Lo¨sung eines Transportproblems. Betonwerk S Fertigteil-Technik 1991, No. 10, pp. 45–50. [49] Hosser, D., Richter, E.: Brandschutzbemessung von Stahlbetonstu¨tzen. Betonwerk S Fertigteil-Technik 2007, pp. 109–113. [50] Ko¨nig, G., Liphardt, S.: Hochhaüser aus Stahlbeton. Beton-Kalender 2003, Teil 1, pp. 1–66. [51] Hock, B., Scha¨fer, K., Schlaich, J.: Fugen und Aussteifungen in Stahlbetonskelettbauten. DAfSt, No. 368, 1986. [52] Schlaich, J., Schober, H.: Fugen im Hochbau – wann und wo? Der Architekt 1976, No. 4. [53] Boll, K.: Anordnung von Dehnfugen bei tragenden Skeletten des Hochbaus. Die Bautechnik 1974, No. 3, pp. 94–98. [54] Stoffregen, U., Ko¨nig, G.: Schiefstellung von Stu¨tzen in vorgefertigten Skelettbauten. Beton- und Stahlbetonbau 1979, No. l, pp. 1–5. [55] Mann, W.: Erdbebenbeanspruchung von Tragwerken und Auswirkung auf Fertigteilkonstruktionen. Fertigteilforum 1981, No. 11, pp. 9–14. [56] Eibl, J., Haüssler-Combe, U.: Baudynamik. Beton-Kalender 1997, Teil 2, pp. 755ff. [57] Bachmann, H.: Erdbebensicherung von Bauwerken, 2nd ed. Birkhaüser Verlag, Basel, 2002.

References

247

[58] Scha¨fer, H.: Die Berechnung von Hochhaüsern als raümlicher Verband von Scheiben, Kernen und Rahmen. Dissertation, Darmstadt TH, 1969. [59] Beck, H., Scha¨fer, H.: Die Berechnung von Hochhaüsern durch Zusammenfassung aller aussteifenden Bauteile zu einem Balken. Bauingenieur 1969, No. 3, pp. 80–87. [60] Schrefler, B.: Zur Berechnung aussteifender Systeme allgemeiner Art von Hochhaüsern. Beton- und Stahlbetonbau 1971, No. 9, pp. 145–153. [61] Johannsen. K.: Berechnung der Aussteifungssysteme von Geschoßbauten mit Tra¨gerrostprogrammen. Beton- und Stahlbetonbau 1976. No. 2, pp. 47–50. [62] Rosman, R.: Die statische Berechnung von Hochhauswa¨nden mit ffnungsreihen. Wilhelm Ernst & Sohn, Berlin, 1965. [63] Sassenberg, H.: Riegelbiegesteifigkeit in gegliederten Scheiben. Bauingenieur 1982, No. l, pp. 17–18. [64] Becker, G.: Die Berechnung gegliederter Hohlka¨sten. Dissertation, Darmstadt TH, 1974. [65] Franzmann, G., Liphardt, S.: Programm HAUS Benutzerhandbuch. KfK-CAD report, Kernforschungszentrum Karlsruhe GmbH. [66] Leonhardt, F., Cziesielski, E.: Beitrag zur Berechnung von Wa¨nden des Großtafelbaus. Die Bautechnik 1967, No. 9, pp. 314–316. [67] Schwing, H. :Wand- und Deckenscheiben aus Fertigteilen. Betonwerk S Fertigteil-Technik 1980, No. 5, pp. 296–301; No. 6, pp. 375–382. [68] Lachmann, H.: Der Einfluß von Fundamentverdrehungen auf die Stabilita¨t (“Labilita¨tszahl”) von Hochbauten. Beton- und Stahlbetonbau 1983, No. 8, pp. 216–217. [69] Brandt, B.: Zur Beurteilung der Gebaüdestabilita¨t nach DIN 1045. Beton- und Stahlbetonbau 1976, No. 7, pp. 177–178. [70] Brandt, B., Schwing, H.: Kriterien zur Vernachla¨ssigung des Windnachweises. Beton- und Stahlbetonbau 1977, No. 3, pp. 53–59. [71] Kordina, K., Quast, U.: Bemessung von schlanken Bauteilen, Knicksicherheitsnachweis. Beton-Kalender 1994, Teil I, pp. 493–587. [72] Graubner, C. A., Scholz, U.: Zum Knicksicherheitsnachweis schlanker Stahlbetondruckglieder mit Knickla¨ngenbeiwert b i 2. Beton- und Stahlbetonbau 1986, No. 8, pp. 217–220. [73] Mehlhorn, G., Schwing, H.: Tragverhalten von aus Fertigteilen zusammengesetzten Scheiben. Research report, Institute for Solid Construction, Darmstadt TH, 1976, No. 33. [74] Stupre Study Group for Building with Precast Concrete Components, Netherlands. Kraftschlu¨ssige Verbindungen im Fertigteilbau, Deckenverbindungen. Beton-Verlag, 1981. [75] Schlaich, J, Scha¨fer, K.: Konstruieren im Stahlbetonbau. Beton-Kalender 2001, Teil II, pp. 311ff. [76] Mehlhorn, G., Schwing, H., Klein, D.: Deckenscheiben aus Bimsbetonhohldielen. Beton- und Stahlbetonbau 1976, No. 6, pp. 142–148. [77] Sto¨ffler, J., Abraham, R.: Aussteifung von Industriebauten mit vorgefertigten Scheiben. Betonwerk S Fertigteil-Technik 1983, No. 13, pp. 5–10. [78] Kreitmann, E., Thoma, F.,Wolff, R.: Neubau eines Gebaüdes fu¨r eine Papier- und Streichmaschine in Hagen-Kabel. Beton- und Stahlbetonbau 1982, No. l, pp. 23–27. [79] Schulz, W., Kaufmann, K.-G.: Bau einer Papiermaschinenhalle. Beton- und Stahlbetonbau 1983, No. 2, pp. 40–44. [80] Fischer, G.: Entwurf und Konstruktion von Hallen fu¨r Papierfabrikationsanlagen. Beton S FertigteilTechnik 1989, No. 5, pp. 57–63. [81] Avramidis, I. E.: Zur Kritik des a¨quivalenten Rahmenmodells fu¨r Wandscheiben und Hochhauskerne. Bautechnik 1991, No. 8, pp. 275–285. [82] Hogeslag, A. J., Stramm, J. P.: Discrete Connections. FIP Symposium, Budapest, 1992, pp. 505–513. [83] Koncz, T.: Ein Vierteljahrhundert Planung und Ausfu¨hrung konstruktiver Bauteile. Betonwerk S Fertigteil-Technik 1985, No. 9, pp. 600–609.

248

5 References

[84] Ro¨sel, W., Sto¨ffler, J.: Beton-Fertigteile im Skelettbau. Beton S Fertigteil-Jahrbuch 1984, pp. 30–65. [85] Schmalhofer, O.: Planerische und statischkonstruktive Grundlagen fu¨r das Bauen mit Stahlbetonfertigteilen. Betonwerk S Fertigteil-Technik 1982, No. 12, pp. 2–6. [86] Gru¨neis, H.: Betonfertigteile als konstruktive und gestalterische Elemente. Betonwerk S FertigteilTechnik. fertigteilforum 1987, No. 19, pp. 8–12. [87] Schwerm, D.: Bauen mit Stahlbetonfertigteilen. BBauBl 1982, No. 12, pp. 851–857. [88] Steinle, A.: Maßkonfektion fu¨r Geschoßbauten. Beispiele von Industriebauten im 6M-System. Zu¨blinRundschau 1976, No. 7/8, pp. 9–12. [89] van Acker, A.: Bemessung von Spannbeton-Hohlplatten und Plattendecken. Betonwerk S FertigteilTechnik 1986, No. l, pp. 35–38. [90] FIP: Design of Multistorey Precast Concrete Structures. FIP Recommendations, Thomas Telford, London, 1986. [91] Landauer, B., Hiller, M.: Herstellung von Stahlbetonfertigteilen fu¨r die King-Saud-Universita¨t in Riyadh/Saudi Arabien. Zu¨blin-Rundschau 1982, No. 14, pp. 20–25. [92] Steinle, A.: CAD/CAM im Massivbau-Rechnerunterstu¨tzung bei Planung und Herstellung der Stahlbetonfertigteile fu¨r die Universita¨t Riyadh. Beton- und Stahlbetonbau 1983, No. 7, pp. 190–196. [93] Rudersdorf, F. A.: Schleuderbeton-Rundstu¨tzen, Gestaltungselemente fu¨r die Architektur als vorgefertigtes Bauteil ho¨chster Qualita¨t. Betonwerk S Fertigteil-Technik 1989, No. 12, pp. 38–43. [94] Stahlbeton-Hohlplatten nach DIN 1045-1. DIBt-Mitteilungen 36 (2005), No. 3. [95] Schnell J., Ackermann F., Nitsch A.: Tragfa¨higkeit von Spannbeton-Fertigdecken auf biegeweichen Auflagern. Beton- und Stahlbetonbau 102 (2007), No. 7, pp. 456ff. [95-1] Hegger, J.: Bemessung und Konstruktion von vorgespannten Decken im Hochbau. Der Pru¨fingenieur, Oct 2003. [96] Schießl, P.: MONTAQUICK – Entwicklung und Anwendung einer montagesteifen Fertigplatte. Betonwerk S Fertigteiltechnik 1981, No. 4, pp. 214–216, No. 5, pp. 291–296. [97] Hahn, V., Steinle, A.: 6 M-System. Bauen S Wohnen 1972, No. 5, pp. 233–236. [98] Schwerm, D.: Bauaufgabe Wand – gelo¨st mit großformatigen Betonfertigteilen. Betonwerk S FertigteilTechnik 1979, No. 9, pp. 515–525. [99] Steinle, A.: Zum Tragverhalten von Blockfundamenten fu¨r Stahlbetonfertigteilstu¨tzen. DBV presentation, Betontag 1981, pp. 186–205. [100] Schwarzkopf, M.: Elementdecke – das Deckensystem mit einem umfassenden Einsatzbereich. Betonwerk S Fertigteil-Technik 1993, No. 6, pp. 54–61. [101] Bechert, H., Furche, J.: Bemessung von Elementdecken mit der Methode der Finite Elemente. Betonwerk S Fertigteil-Technik 1993, No. 5, pp. 47–51. [102] Goldberg, G., Schmilz, M., Land, H.: Zweiachsige Lastabtragung bei Elementdecken. Betonwerk S Fertigteil-Technik 1993, No. 7, pp. 86–89. [103] Land, H.: Teilfertigdecken. Betonwerk S Fertigteil-Technik 1994, No. 5, pp. 93–95, No. 6, pp. 108–118. [104] Mu¨hlbauer, S., Stenzel, G.: Kompaktstu¨tzen aus hochfestem Beton, Konstruktion und Bemessung. Beton- und Stahlbetonbau 98 (2003), No. 12, pp. 678–686. [105] Weiske, R.: Durchleitung hoher Stu¨tzenlasten bei Stahlbeton-Flachdecken. Dissertation, Braunschweig TU, Massivbau und Brandschutz, No. 180, 2004. [106] Falkner, H., Eierle, B., Henke, V.: HH-Stu¨tzen – schlanke Betonfertigteile aus Hochleistungsbeton. Beton S Fertigteil Jahrbuch 2003, pp. 130–139. [107] Forster, J.: Doppelwandplatten – die innovative Lo¨sung. Beton S Fertigteil-Jahrbuch, 2003, pp. 208–212. [108] DAfStb: Richtlinie Wasserundurchla¨ssige Bauwerke aus Beton, 2003, und Berichtigung zur WU-Richtlinie, 2006. [109] DAfStb: Erlaüterungen zur DAfStb-Richtlinie Wasserundurchla¨ssige Bauwerke aus Beton, 2006.

References

249

[110] Hohmann, R.: Elementwa¨nde in dru¨ckendem Grundwasser – Diskrepanz zwischen Theorie und Praxis? Beton- und Stahlbetonbau 102 (2007), No. 12. [111] Schmalhofer, O.: Fassaden aus Stahlbetonfertigteilen. Bauingenieur 1985, pp. 211–216. [112] Strumpf, K.: Neue Entwicklungen der vorgefertigten Stahlbetonfassade. DBZ 1983, No. l, pp. 77–81. [113] Haferland, F.: Das Wa¨rme-, Diffusions- und schalltechnische Verhalten von Beton-Außenwa¨nden. Betonfertigteilforum 11/1971. Betonstein-Zeitung 1971, No. 3, pp. 3–22. [114] Energieeinsparverordnung (Energy Conservation Act), EnEV 2007. [115] Middel, M.: Die Energieeinsparverordnung: Chancen fu¨r die Betonbauweise im Wohnungsbau. Beton S Fertigteil-Jahrbuch 2003, pp. 172–180. [116] Brandt, J., Krieger, R., Moritz, H.: Außenwa¨nde aus Betonbauteilen: Wa¨rme-, Schall- und Brandschutz. Beton S Fertigteil-Jahrbuch 1983. [117] Cziesielski, E., Raabe, B.: Tauwasserschutz. Bauphysik 1985 6/82, 2/83, 1/84, 4/84, 3/85, 4/85, 6/85. [118] Recknagel, Sprenger: Taschenbuch fu¨r Heizung und Klimatechnik. Oldenbourg Verlag, Munich, 1986. [119] PCI: Architectural Precast Concrete (various brochures covering: design, weathering, cladding, loadbearing walls, joint details, etc.), PCI, Chicago. [120] Hofbauer, H.: Bauphysikalische bauspezifische Problemstellungen und Vorschla¨ge zu ihrer Konstruktive Lo¨sung. Betonwerk S Fertigteil-Technik 1982, No. 12, pp. 7–14. [121] Huberty, J. M.: Fassaden in der Witterung. Beton-Verlag 1983. [122] Cziesielski, E., Kotz, D.: Betonsandwich-Wa¨nde, Bemessung der Vorsatzschalen und Ausbildung der Fugen. Beton S Fertigteil-Jahrbuch 1984, pp. 66–122. [123] Cziesielski, E.: Belu¨ftete Fugen. Betonwerk S Fertigteil-Technik 1978. No. 8, pp. 441–447. [124] Martin, B.: Joints in Buildings, John Wiley & Sons, 1977 [125] Stiller, M.: Das Verarbeiten von Fugenmassen im Betonfertigteilbau. Beton- und Stahlbetonbau 1968, No. 7, pp. 156–159. [126] Steinle, A.: Das Zu¨blin-Haus. Betonwerk S Fertigteil-Technik 1985. No. 6. pp. 374–383. [127] Halfen, facade fixings. Company Brochure 11/2006. [128] Deha: Verbundanker-Systeme, Fassadenankersysteme, Transportanker-Systeme. Company brochure 1996. [129] Cziesielski, E., Kotz, D.: Temperaturbeanspruchung mehrschichtiger Stahlbetonwa¨nde. Betonwerk S Fertigteil-Technik 1984, No. l, pp. 28–29. [130] Utescher G.: Der Tragsicherheitsnachweis fu¨r dreischalige Außenwandplatten (Sandwichplatten) aus Stahlbeton. Die Bautechnik 1973, No. 5, pp. 163–171. [131] Utescher, G.: Beurteilungsgrundlagen fu¨r Fassadenverankerungen. Wilhelm Ernst & Sohn, Berlin, 1978. [132] Krieger, R.: Bauphysikalische Fragen bei Betonfertigteil-Wandtafeln. Betonwerk S Fertigteil-Technik 1983, No. 2, pp. 84–89. [133] Huyghe, G.: Verankerungen und Transportanker fu¨r Fertigbauteile. B1BM 84, 1984, pp. 650–686. [134] Taylor, H. P. J.: Precast Concrete Cladding. Edward Arnold, London, 1992. [135] Lutz, Lenisch, Klopfer, Freymuth, Krampf: Lehrbuch der Bauphysik, 2nd ed. Teubner Verlag, Stuttgart, 1989. [136] Mehl, R.: Außerirdisch intelligent – Das Phaeno in Wolfsburg. Beton S Fertigteil-Jahrbuch 2007, pp. 8ff. [137] Mehl, R.: Haus hinter Gittern – Laborgebaüde in Wageningen. Beton S Fertigteil-Jahrbuch 2008, pp. 8ff. [138] Mehl, R.: Die Scho¨pfung und das Licht – Gemeindezentrum in Mannheim-Neuhermsheim. Beton S Fertigteil-Jahrbuch 2008, pp. 32–37. [139] Hegger, J., Will, N., Schneider, H. N., Scha¨tzke, C., Curbach, M., Jesse, F.: Fassaden aus textilbewehrtem Beton. Beton S Fertigteil-Jahrbuch 2005, pp. 76–82.

250

5 References

[140] National Technical Approval No. Z-33.1-577: Fassadenplatten aus Betonwerkstein mit ru¨ckseitig einbetonierten Befestigungselementen zur Verwendung bei hinterlu¨fteten Fassaden. DIBt 2004. [141] Curbach, M. Speck, K.: Lasteinleitung in du¨nnwandige Bauteile aus textilbewehrtem Beton mit kleinen Du¨beln. Abschlussbericht zum Forschungsvorhaben V 426 des DAfStb, Institute for Solid Construction, Dresden TU, 2003. [142] Lukas, B.: “Bodensee Kiesel” – Ein Veranstaltungssaal in Friedrichshafen. Beton S Fertigteil-Jahrbuch 2008, pp. 70–74. [143] DIN 1045-1: Concrete, reinforced and prestressed concrete structures – Part 1: Design construction. Jul 2007 [144] DIN 1045-1/A1: Concrete, reinforced and prestressed concrete structures – Part 1: Design construction. Amendment. Jan 2008 [145] Tillmann, M.: Verbundfugen im Fertigteilbau – Entwicklung der technischen Regelwerke. Beton- S Fertigteil-Jahrbuch 2008. [146] Zilch, K.; Mu¨ller, A.: Grundlagen und Anwendungsregeln der Bemessung von Fugen nach EN 1992-1-1. Final report, DIBt research project, Chair of Solid Construction, Munich TU, Apr 2007. [147] DAfSt, No. 525, Erlaüterungen zu DIN 1045-1, Berlin, 2003. [148] National Technical Approval No. Z-15.6-204: Halfen HDB-E-Anker zur Verankerung in Rahmenendknoten und Konsolen. DIBt, Berlin, 2002. [149] Daschner, F., Kupfer, H.: Literaturstudie zur Schubsicherung bei nachtra¨glich erga¨nzten Querschnitten. DAfSt 1986, No. 372. [150] Daschner, F.: Versuche zur notwendigen Schubbewehrung zwischen Betonfertigteilen und Ortbeton. DAfSt 1986, No. 372. [151] Nissen, L, Daschner, F., Kupfer, H.: Verminderte Schubdeckung in Stahlbeton- und Spannbetontra¨gern mit Fugen parallel zur Tra¨gerrichtung und nicht vorwiegend ruhender Belastung. DAfSt 1986, No. 372. [152] Rehm, G., Eligehausen, R., Paul: Verbund in Fugen von Platten ohne Schubbewehrung. Kurzberichte aus der Bauforschung 1980, 3/80–36, pp. 191–194. [153] Daschner, F., Kupfer, H.: Durchlaufende Deckenkonstruktionen aus Spannbetonfertigteilen mit erga¨nzender Ortbetonschicht. Betonwerk S Fertigteil-Technik 1983, No. 11, pp. 714–721. [154] FIP: Shear at the interface of precast and in situ concrete. FIP, Wexham Springs, 1982. [155] Suikka, A.: Verbund-Konstruktionen mit Hohlplatten. Betonwerk S Fertigteil-Technik 1986, No. 5, pp. 315–316. [156] Steinle, A.: Zur Frage der Mindestabmessungen von Konsolen. Beton- und Stahlbetonbau 1975, No. 6, pp. 150–153. [157] Steinle, A., Rostasy, F. S.: Zum Tragverhalten ausgeklinkter Tra¨gerenden. Betonwerk S Fertigteil-Technik 1975, No. 6, pp. 270–277. [158] Steinle, A.: Zum Tragverhalten ausgeklinkter Tra¨gerenden. Presentation, Betontag 1975, DBV e.V. 1975, pp. 364–376. [159] Graubner, C.-A.: Zur Bemessung von Stahlbetonbalken bei unsymmetrischer Belastung aus Konsolba¨ndern. Bauingenieur 1984, vol. 59, pp. 221–223. [160] Raths, Charles H.: Spandrel Beam Behavior and Design, 1983, pp. 62–130. [161] Deneke, O., Holz, K., Litzner, H.-U.: bersicht u¨ber praktische Verfahren zum Nachweis der Kippsicherheit schlanker Stahlbeton- und Spannbetontra¨ger. Beton- und Stahlbetonbau 1985, No. 9, pp. 238–243; No. 10, pp. 274–280; No. 11, pp. 299–304. [162] Zilch, K., Staller A., Johring A.: Vergleichende Untersuchungen zum Tragsicherheitsnachweis kippgefa¨hrdeter Stahlbeton- u. Spannbetontra¨ger nach Theorie 2. Ordnung. Bauingenieur 1997, pp. 157–165. [163] Rafla, K.: Na¨herungsverfahren zur Berechnung der Kipplasten von Tra¨gern mit in La¨ngsrichtung beliebig vera¨nderlichem Querschnitt. Die Bautechnik 1975, No. 8, pp. 269–275. [164] Stiglat, K.: Die Kippsicherheit von Beton- und Stahlbetonbalken. Presentation, Checking Engineers Conference, Freudenstadt, 1971.

References

251

[165] Klang, H.: Einfluß der Elastizita¨t der Aufha¨ngevorrichtung auf die Kippstabilita¨t des an zwei Punkten aufgeha¨ngten Tra¨gers. Beton- und Stahlbetonbau 1965, No. 11, pp. 271–273. [166] Stiglat, K.: Na¨herungsberechnung der kritischen Kipplasten von Stahlbetonbalken. Bautechnik 1971, No. 3, pp. 98–100. [167] Dilger, W.: Vera¨nderlichkeit der Biege- und Schubfestigkeit bei Stahlbetontragwerken und ihr Einfluß auf Schnittkraftverteilung und Traglast bei statisch unbestimmter Lagerung. DAfStb 1966, No. 179. [168] Mann, W.: Kippnachweis und Kippaussteifung von schlanken Stahlbeton- und Spannbetontra¨gern. Beton- und Stahlbetonbau 1976, No. 2, pp. 37–42. [169] Mann, W.: Anwendung des vereinfachten Kippnachweises auf T-Profile aus Stahlbeton. Beton- und Stahlbetonbau 1985, No. 9, pp. 235–243. [170] Streit, W., Mang, R.: berschla¨glicher Kippsicherheitsnachweis fu¨r Stahlbeton- und Spannbetonbinder. Bauingenieur 1984, pp. 433–439, also 1985, pp. 368. [171] Streit, W., Gottschalk, H.: berschla¨gige Bemessung von Kipphalterungen fu¨r Stahlbeton- und Spannbetonbinder. Bauingenieur 1986, pp. 555–559. [172] Dieterle, H., Steinle, A.: Blockfundamente fu¨r Stahlbetonfertigstu¨tzen. DAfStb 1981, No. 326. [173] Kordina, K., No¨lting, D.: Tragfa¨higkeit durchstanzgefa¨hrdeter Stahlbetonplatten. DAfSt, 1986, No. 371. [174] Holz, K.: Baulicher Brandschutz mit Beton (Teil 4). Brandschutztechnische Bemessung waagerechter Bauteile (Balken und Platten, etc.). Betonwerk S Fertigteil-Technik 1982, No. 11, pp. 676–681. [175] Meyer-Ottens, C: Baulicher Brandschutz mit Beton (2. Teil); Fugen, Lager und Sonderbauteile. Betonwerk S Fertigteil-Technik 1982, No. 9, pp. 555–559. [176] Mehlhorn, G., Ro¨der, F.-K., Schulze, J. U.: Zur Kippstabilita¨t vorgespannter und nicht vorgespannter, parallelgurtiger Stahlbetontra¨ger mit einfach symmetrischem Querschnitt. Beton- und Stahlbetonbau 1991, No. 2, pp. 25–32; No. 3, pp. 59–64. [177] Ro¨der, F.-K.: Ermittlung wirklichkeitsnaher Querschnittswerte u. Steifigkeiten fu¨r vorgespannte und nicht vorgespannte Rechteck- und T-Querschnitte aus Stahlbeton. Beton- und Stahlbeton 1990, No. 6, pp. 154–159; No. 7, pp. 180–185. [178] Kraus, D., Ehret, K.-H.: Berechnung kippgefa¨hrdeter Stahlbeton- und Spannbetontra¨ger nach der Theorie II. Ordnung. Beton- und Stahlbetonbau 1992, No. 5, pp. 113–118. [179] Ko¨nig, G., Pauli, W.: Ergebnisse von sechs Kippversuchen an schlanken Fertigteiltra¨gern aus Stahlbeton und Spannbeton. Beton- und Stahlbetonbau 1990, No. 10, pp. 253–258. [180] Ko¨nig, G., Pauli, W.: Nachweis der Kippstabilita¨t von schlanken Fertigteiltra¨gern aus Stahlbeton und Spannbeton. Beton- und Stahlbetonbau 1992, No. 5, pp. 109-112; No. 6, pp. 149–151. [181] Stiglat, K.: Kippnachweis bei niedrigen Vergleichsschlankheiten lv. Beton- und Stahlbeton 1996, No. 12, pp. 292. [182] Bacher, W.: berpru¨fung der Gu¨te eines preisgerechten Na¨herungsverfahrens zum Nachweis der Kippsicherheit schlanker Stahlbeton- und Spannbetontra¨ger. Beton- und Stahlbeton 1995, No. 7, pp. 176– 179, No. 9, pp. 209–213. [183] Winner, H.: Kippsicherheit u¨ber Vergleichsschlankheit in Rahmen im Eurocode 2. Beton- und Stahlbeton 1998, No. 1, pp. 20–22. [184] Stiglat, K.: Zur Na¨herungsbetrachtung der Kipplasten von Stahlbeton- und Spannbetontra¨gern u¨ber Vergleichsschlankheiten. Beton- und Stahlbetonbau 1991, No. 10, pp. 237–240; 1996 No. 12, pp. 292. [185] Ru¨der, F.-U.: Ein Na¨herungsverfahren zur Beurteilung der Kippstabilita¨t von Satteldachbindern aus Stahlbeton oder Spannbeton. Beton- und Stahlbeton 1997, No. 11, pp. 301–307; No. 12, pp. 341–347. [186] Mattheiß, J.: Abscha¨tzung einer sicheren Druckflanschbreite. Beton- und Stahlbeton 1999, No. 7, pp. 289–294. [187] Schaller, G.: Einfluss der Gabelsteifigkeit auf das Kippverhalten- Bemessungsmoment der Gabel. Beton- und Stahlbetonbau 1997, pp. 73–78. [188] Haß, R., Wesche, J.: Bemessung feuerbesta¨ndiger gegliederter Außenwandelemente aus Stahlbetonfertigteilen (“Lochfassaden”). Betonwerk S Fertigteil-Technik 1988, No. 8, pp. 24–28.

252

5 References

[189] Mainka, G.-W., Paschen, H.: Untersuchungen u¨ber das Tragverhalten von Ko¨cherfundamenten. DAfStb, No. 411, 1990. [190] Eligehausen, R., Gerster, R.: Das Bewehren von Stahlbetonbauteilen. Erlaüterungen zu verschiedenen gebraüchlichen Bauteilen. DAfStb, No. 399, 1993. [191] Ackermann, G., Burckhardt, M.: Tragverhalten von bewehrten Verbundfugen bei Fertigteilen und Ortbeton in den Grenzzusta¨nden der Tragfa¨higkeit und Gebrauchstauglichkeit. Beton- und Stahlbetonbau 1992, No. 7, pp. 165–170; No. 8, pp. 197–200. [192] Kordina, K.: Zur Berechnung und Bemessung von Einzel-Fundamentplatten nach EC2 Teil 1. Betonund Stahlbetonbau 1994, No. 6, pp. 224–226. [193] Eibl, J., Zeller, W.: Untersuchungen zur Traglast der Druckdiagonalen in Konsolen. Beton- und Stahlbetonbau 88 (1993), No. 1, pp. 23–26. [194] Reineck, K.-H.: Modellierung der D-Bereiche bei Fertigteilen. Beton-Kalender 2005, Teil 2, pp. 241ff. [195] Stenzel, G; Fingerloos, F.: Konstruktion und Bemessung von Details nach DIN 1045-1. Beton-Kalender 2007, Teil 2. [196] Hegger, J; Roeser, W.; Lotze, D. Kurze Verankerungsla¨ngen mit Rechteckankern – Bewehrung S Bauausfu¨hrung. Beton- und Stahlbetonbau (2004), No. 1, pp. 1–9. [197] Graubner, C.-A., Hausmann, G., Karasek, J.: Bemessung von Betonfertigteilen nach DIN 1045-1. Beton-Kalender 2005, Teil 2. [198] Fachvereinigung Deutscher Betonfertigteilbau eV.: fire protection leaflet, 04/2008 (draft). [199] Nause, P.: Brandverhalten von hochfestem Beton. Beton S Fertigteil-Jahrbuch 2003, pp. 150ff. [200] Verband der Sachversicherer: Merkblatt fu¨r die Anordnung und Ausfu¨hrung von Brand- und Komplextrennwa¨nden. VDS Verlag 2234, Jul 2006. [201] DBV leaflet “Abstandhalter”, Jul 2002. [202] Stupre Study Group for Building with Precast Concrete Components, Netherlands. Kraftschlu¨ssige Verbindungen im Fertigteilbau. Konstruktions-Atlas, 2nd ed., Beton-Verlag, 1987. [203] Hahn, V., Hornung, K.: Untersuchungen von Mo¨rtelfugen unter vorgefertigten Stahlbetonstu¨tzen. Betonsteinzeitung 1968, No. 11, pp. 353–362. [204] Suter, R.: Mo¨rtelverbindungen. Betonwerk S Fertigteil-Technik 1975, No. 11, pp. 531–536. [205] Brandt, B., Scha¨fer, H. G.: Verbindung von Stahlbetonfertigteilstu¨tzen. Forschungsreihe der Bauindustrie 1974, vol. 18. [206] Paschen, H., Zillich, V.: Der Stumpfstoß von Fertigteilstu¨tzen. Betonwerk S Fertigteil-Technik 1980, No. 5, pp. 279–285; No. 6, pp. 360–364. [207] Paschen, H., Stockleben, U., Zillich, V.: Querzugbeanspruchung durch Mo¨rtelfugen infolge Mo¨rtelquerdehnung und Teilfla¨chenbelastung. Betonwerk S Fertigteil-Technik 1981, No. 7, pp. 385–392. [208] Rahlwes, K.: Lagerung und Lager von Bauwerken. Beton-Kalender 1995, Teil II, pp. 631. [209] Ko¨nig, G., Tue, N. V., Saleh, H., Kliver, J.: Herstellung und Bemessung stumpf gestoßener Fertigteilstu¨tzen. Beton S Fertigteil-Jahrbuch, 2003. [210] Saleh, H.: Ein Beitrag zur Untersuchung und Bemessung von stumpf gestoßenen Fertigteilstu¨tzen aus normalfestem Beton. Dissertation, University of Leipzig, 2002. [211] Ko¨nig, G., Minnert, J.: Tragverhalten von stumpf gestoßenen Fertigteilstu¨tzen aus hochfestem Beton. Beton S Fertigteil-Jahrbuch 2000, pp. 81–94. [212] Ko¨nig, G., Minnert, J., Saleh, H.: Stumpf gestoßene Fertigteilstu¨tzen aus Normalbeton. Beton S Fertigteil-Jahrbuch 2001, pp. 110–121. [213] Maurer, Breitbach, Jhouahra: Neues Nachweiskonzept fu¨r die Bemessung von Lagern und Lagerungen im Fertigteilbau nach DIN EN 1337. Betonwerk S Fertigteil-Technik 02 (2007), pp. 50–51. [214] Eggert, H.: Lager im Bauwesen. Anmerkungen zur Herausgabe der Normenreihe DIN 4141. Beton- und Stahlbetonbau 1988, No. 7, pp. 193–198.

References

253

[215] Vinje, L.: Bemessung von unbewehrten Elastomerlagern in Betonfertigteilbauten. Betonwerk S Fertigteil-Technik 1985, No. 5, pp. 306–314; No. 6, pp. 392–398. [216] Mu¨ller, F., Sasse, R., Thorma¨hlen, U.: Stu¨tzensto¨ße im Stahlbeton-Fertigteilbau mit bewehrten Elastomerlagern. DAfStb & Beton- und Stahlbetonbau 1982, No. 11/12; 1982, No. 339. [217] Kessler, E., Schwerm, D.: Unebenheiten und Schiefwinkligkeiten der Auflagerfla¨chen fu¨r ElastomerLager bei Stahlbetonfertigteilen. Betonwerk S Fertigteil-Technik 1983, No. 13, pp. 1–5. [218] Kordina, K., No¨lting, D.: Zur Auflagerung von Stahlbetonteilen mittels unbewehrten Elastomerlagern. Bauingenieur 1981, pp. 41–44. [219] Flohrer, M., Stephan, E.: Bemessungsdiagramm fu¨r die Querzugkra¨fte bei Elastomerlagern. Die Bautechnik 1975, No. 9, pp. 296–301; No. 12, pp. 420–427. [220] Battermann, W.: Mo¨glichkeiten der Einflussnahme auf die Eigenschaften von Elastomerlagern zur funktionsgerechten Auflagerung von Betonbauteilen. Betonwerk S Fertigteil-Technik 1978, No. l, pp. 37–41. [221] Vinje, L.: Auflager-Zwischenschichten fu¨r Spannbeton-Hohlplatten. Betonwerk S Fertigteil-Technik 1986, No. 10, pp. 636–641. [222] Sasse, H.-R.: Gleit- und Verformungslager im Hoch- und Bru¨ckenbau. VDI Reports 1980, No. 384. [223] Kessler, E.: Die Anwendung unbewehrter Elastomerlager. Betonwerk S Fertigteil-Technik 1987, No. 8, pp. 419–429. [224] Vambersky, J. N. J. A., Walraven, J. C.: Die Tragfa¨higkeit von auf Druck beanspruchten unbewehrten Mo¨rtelfugen. Betonwerk S Fertigteil-Technik 1988, No. 7, pp. 66–73. [225] Paschen, H.: Untersuchung von durch Zwang einachsig exzentrisch belasteten Stu¨tzensto¨ßen des Betonfertigteilbaus. Bauingenieur 1992, pp. 69–76. [226] Hasse, E.: Zum Tragfa¨higkeitsnachweis fu¨r Wand-Decken-Knoten im Großtafelbau. DAfStb 1982, No. 328. [227] Franke, H.: Die Schweißverbindungen in Stahlbetonbauteilen. Konstruktiver Ingenieurbau, VBI 1986. [228] CEN-TS: Design of Fastenings for Use in Concrete – Final Draft, 2005. [229] Czychy,: Zuku¨nftige Bemessung der Ankerschienen. Betonwerk S Fertigteil-Technik 04/2008. [230] Rehm, G., Eligehausen, R., Mallee, R.: Befestigungstechnik. Beton-Kalender 1997, Teil II, pp. 609ff. [231] Bertram, D.: Betonstahl, Verbindungselemente, Spannstahl. Beton-Kalender 2002, Teil 1, pp. 153ff. [232] Cziesielski, E., Utescher, G., Paschen, H.: Bemessungsvorschla¨ge fu¨r Bolzenverbindungen. DAfStb 1983, No. 346. [233] Wiedenroth, M.: Einspanntiefe und zula¨ssige Belastung eines in einen Betonko¨rper eingespannten Stabes. Die Bautechnik 1971, No. 12, pp. 426–429. [234] Utescher, G., Herrmann, H.: Befestigungs- und Verbindungsmittel beim Beton S Fertigteilbau. Kurzbericht aus der Bauforschung 1978, Nr. II. [235] Reißmann, K.: ber die Bemessung von Transportankern fu¨r Stahlbetonfertigteile. Betonwerk S Fertigteil-Technik 1982, No. 7, pp. 406–409. [236] Eibl, J., Schu¨rmann, U.: HV-Schraubenanschlu¨sse fu¨r Stahlbetonkonsolen. Bauingenieur 1982, pp. 61–68. [237] Zu¨blin: Versuche mit Stahlkonsolen. Unpublished report, 1985. [238] Eligehausen, R., Asmus, J., Lotze, D., Potthoff, M.: Ankerschienen. Beton-Kalender 2007, pp. 375ff. [239] Randl, N.: Tragverhalten einbetonierter Scherbolzen. Beton- und Stahlbetonbau 100 (2005), No. 6, pp. 467–474. [240] Asmus, J., Eligehausen, R., Schneider, J.: Schubdorne nach neuer DIN 1045-1. Betonwerk S FertigteilTechnik (2008), No. 2, pp. 108–109. [241] VDI Directive: Bemessung und Anwendung von Transportankersystemen fu¨r Betonfertigteile (draft 05/ 2007).

254

5 References

[242] Mattock, A. H., Johal, L., Chow, H. C: Shear transfer in reinforced concrete with moment or tension acting across the shear plane. PCI-Journal 1975, No. 7/8, pp. 76–93. [243] Mehlhorn, G., Schwing, H.: Zur Berechnung und Konstruktion von Wandscheiben aus Fertigteilen. Betonwerk S Fertigteil-Technik 1973, No. 5, pp. 360–370. [244] Paschen, H., Zillich, V. C.: Tragfa¨higkeit querkraftschlu¨ssiger Fugen zwischen Stahlbeton-Fertigteildecken. DAfStb 1983, No. 348. [245] Paschen, H., Zillich, V. H.: Tragfa¨higkeit querkraftschlu¨ssiger Fugen zwischen vorgefertigten Stahlbeton-Fertigteildecken. Beton- und Stahlbetonbau 1983, No. 6, pp. 168–172; No. 7, pp. 197–201. [246] Paschen, H.: Berichtigung zu [3.3.5]. Beton- und Stahlbetonbau 1987, No. 2, pp. 56. [247] Griebenow, G., Koch, R., Sitka, R., Vo¨lkel, G.: Tragende Deckenscheiben mit Gasbeton-Fertigteilen. Experimentelle Untersuchungen und Berechnungsmodelle. Betonwerk S Fertigteil-Technik 1989, No. 6, pp. 56–61; No. 8, pp. 62–66. [248] Cholewicki, A.: Schubu¨bertragung bei La¨ngsverbindungen von Hohlplattendecken. Betonwerk S Fertigteil-Technik 1991, No. 4, pp. 58–67. [249] Kreuser, W., Schulenberg, W.: Fugenversuche an Stahlbetonhohlplatten. Beton- und Stahlbetonbau 1989, No. 10, pp. 245–248. [250] Koncz, T.: Anlagen und Formen zur Produktion großformatiger Fertigteile. BIBM 84, presentation, 1984, pp. 603–616. [251] Fogarasi, F. G.: Herstellen von Spannbetonelementen nach dem Umlaufverfahren. Betonwerk S Fertigteil-Technik 1986, No. 5, pp. 326–333. [252] Schwarz, S.: Garagen- und Treppenherstellung im Umlaufverfahren. Betonwerk S Fertigteil-Technik 1982, No. 10, pp. 623–631. [253] Reymann, W.: Transport- und Bewegungsablaüfe in Fertigteilwerken in komplexer Betrachtung. BIBM 84, presentation, 1984, pp. 622–649. [254] Koncz, T.: Anlagen fu¨r den Großtafelbau in verschiedenen Varianten. Betonwerk S Fertigteil-Technik 1983, No. 4, pp. 242–249. [255] Schwarz, S.: Moderne Maschinen und Anlagen zur Herstellung unterschiedlicher Deckenelemente. Betonwerk S Fertigteil-Technik 1985, No. 1, pp. 4–20; No. 2, pp. 96–106. [256] Reymann, W.: Neue Konzepte des Produktionsablaufs in Fertigteilwerken. BIBM 87, presentation,1987, draft, pp. 1–14. [257] Markus, M.: Manufacturing systems for hollow-core concrete slabs, vol. 3, FIP Congress 1982, pp. 244– 261. [258] Schwarz, S.: Realisierte Schritte zur computergesteuerten Betonfertigteilproduktion. Finland: Extruder mit Shear-Compaction-System, CAD/CAM, Betontechnol. Betonwerk S Fertigteil-Technik 1987, No. 3, pp. 152–159. [259] Hoffmann, O.: Fertigungssteuerung und Terminkontrolle in Fertigteilwerken mit Hilfe der EDV. Betonwerk S Fertigteil-Technik 1984, No. 12, pp. 799–806. [260] Kra¨mer, R.: Tammer Elementii – ein nahezu automatisiertes Fertigteilwerk. Betonwerk S Fertigteil-Technik 1992, No. 3, pp. 166–180. [261] Reymann,W.: Das automatische Betonwerk –Wirklichkeit und Vision. Betonwerk S Fertigteil-Technik 1992, No. 4, pp. 88–96. [262] Ehmer, M.: C-Techniken fu¨r die Fertigteilproduktion. Betonwerk S Fertigteil-Technik 1992, No. 4, pp. 106–113. [263] Hohmann, H., Ehlert, W.: Die besonderen Mo¨glichkeiten des CAD-Einsatzes in Fertigteilwerken. Bauinformatik 1990, No. 2, pp. 53–60. [264] Ehlert, W.: CAD/CAM fu¨r Industriebauteile aus der Sicht CAD und PPS. Betonwerk S FertigteilTechnik 1993, No. 12, pp. 87–93. [265] Schwo¨rer, D.: Ansa¨tze zu einer CAD-CAM-Produktion von stabfo¨rmigen Betonfertigteilen bei Schwo¨rer. Betonwerk S Fertigteil-Technik 1993, No. 8, pp. 84–91.

References

255

[266] Strauch, J.: Betriebsdatenerfassung in Betonfertigteilwerken. Betonwerk S Fertigteil-Technik 1991, No. 2, pp. 40–45. [267] Stadimann, B., Weckenmann, H.: Umweltfreundlicher Schalungsroboter vermeidet Styropor und schwere Arbeit. Betonwerk S Fertigteil-Technik 1993, No. 12, pp. 81–87. [268] Widmann, H, Enoekl, V.: Schaumbeton-Baustoffeigenschaften, Herstellung. Betonwerk S FertigteilTechnik 1991, No. 6, pp. 38–44. [269] Tebbe, R.: Anorganische Pigmente, grundsa¨tzliche Eigenschaften und Herstellung. Fachveranstaltung “Eingefa¨rbter Beton”, Haus der Technik, Essen, 1990, presentation. [270] Reinhardt, H.-W.: Beton. Beton-Kalender 2007, Teil 1, pp. 353ff. [271] Heufers, H., Schulze,W.: Neuartige Oberfla¨chengestaltung mit farbigen Zuschla¨gen. Betonwerk S Fertigteil-Technik 1980, No. 9, pp. 531–539. [272] Weigler, H., Karl, S., Jaegermann, C.: Leichtzuschlagbeton mit hohem Gehalt an Mo¨rtelporen. DAfStb 1981, No. 321. [273] Heufers, H.: Leichter Normalbeton. Betonwerk S Fertigteil-Technik 1976, No. 11. [274] Wesche, K.: Baustoffe fu¨r freitragende Bauteile, Teil 2: Beton. Bauverlag, 1981. [275] Nischer, P.: Schaumbeton. Betonwerk S Fertigteil-Technik 1983, No. 3, pp. 148–151. [276] Bodner, H., Nothdurft, K.: Landwirtschaftsschulen und Handwerkliche Berufsschulen im Irak (ASI). Zu¨blin Rundschau 1984, No. 16, pp. 26–32. [277] Clausen, H. P.: Zunehmende Verwendung, von geschoßhohen, raumgroßen Wandtafeln aus haufwerksporigem Leichtbeton. Betonwerk S Fertigteil-Technik 1985, No. 12, pp. 780–786. [278] Hohwiller, F., Ko¨hling, K.: Styropor-Beton. Betonsteinzeitung 1968, No. 2, pp. 81–87; No. 3, pp. 132–137. [279] Sass: Porenleichtbeton mit neopor fu¨r Industrie-, Wohn- und Kommunalbauten. Instruction manual, Gloria-Transportbeton GmbH, Buren. [280] Linder, R.: Stand der Technik bei faserversta¨rktem Beton. Tiefbau 1975, No. 5, pp. 321–330. [281] PCI: Glass Fiber Reinforced Concrete Cladding. Brochure, Prestressed Concrete Institute (PCI), Chicago. [282] Held, M., Ko¨nig, G.: Hochfester Beton bis B 125 – ein geeigneter Baustoff fu¨r hochbelastete Druckglieder. Betonwerk S Fertigteil-Technik 1992, No. 2, pp. 41–45; No. 3, pp. 74–76. [283] Kern, E.: Technologie des hochfesten Betons. Beton 1993, No. 3, pp. 109–115. [284] Plenker, H.-H.: Dosierung und Verteilung von Pigmenten in Beton. Betonwerk S Fertigteil-Technik 1991, No. 9, pp. 58–65. [285] Kresse, P.: Ausblu¨hungen und ihre Verhinderung. Betonwerk S Fertigteil-Technik 1991, No. 10, pp. 73–88. [286] Vambersky, J. N. J. A.: Bemessungsregeln fu¨r architektonische Elemente aus Glasfaserbeton. Betonwerk S Fertigteil-Technik 1989, No. 7, pp. 24–33. [287] Meyer, A.: Konstruktions- und Bemessungsregeln fu¨r Glasfaserbeton. Betonwerk S Fertigteil-Technik 1990, No. 12, pp. 49–53. [288] Meyer, A.: Glasfaserbeton – Baustoff mit Zukunft. Entwicklung, Verfahren, Gera¨te. Beton 1991, No. 6, pp. 277–281. [289] Meyer, A.: Wellcrete – eine fortschrittliche Technologie fu¨r die kostengu¨nstige Produktion hochwertiger Faserbetonprodukte. Betonwerk S Fertigteil-Technik 1991, No. 8, pp. 70–79. [290] Curiger P.: Bemessen von Bauteilen aus Glasfaserbeton. Betonwerk S Fertigteil-Technik 1994, No. 7, pp. 59–67. [291] Bergmeister, K.: Konstruieren mit Fertigteilen. Beton-Kalender 2005, Teil 2, pp. 163ff. [292] Ko¨nig, G., Grimm, R.: Hochleistungsbeton. Beton-Kalender 2000, Teil 2, pp. 327ff. [293] Hillemeier, B., Buchenau, G., Herr, R. et al.: Spezialbetone. Beton-Kalender 2006, Teil 1, pp. 519ff.

256

5 References

[294] Falkner, H., Teutsch, M.: Stahlfaserbeton – Anwendungen und Richtlinie. Beton-Kalender 2006, Teil 1, pp. 665ff. [295] Adam, T., Proske, T.: Hochfester SVB mit hoher Fru¨hfestigkeit zur Herstellung von Betonfertigteilen mit sofortigem Verbund. Mischungszusammensetzung und Untersuchung der bemessungsrelevanten Eigenschaften. Betonwerk S Fertigteil-Technik 2007, No. 12, pp. 12–20. [296] Hubertova, M.: Fertigteile aus selbstverdichtetem Leichtbeton fu¨r das Stadion des SK Slavia Prag. Beton S Fertigteil-Jahrbuch 2007, pp. 22–31. [297] Hartz, U.: Normen und Regelwerke. Beton-Kalender 2005, Teil 2, pp. 502ff. [298] Holschemacher, K., Klug, Y., Dehn, F., Wo¨rner, J.-D.: Faserbeton. Beton-Kalender 2006, Teil 1, pp. 585ff. [299] Ko¨nig, G., Dehn, F., Holschemacher, K.,Weiße, D.: Verbundverhalten von Betonstahl in Hochleistungsleichtbeton unter dynamischer Beanspruchung. Beton S Fertigteil-Jahrbuch 2002, pp. 149–158. [300] Falkner, H., Teutsch, M.: Entwicklung duktiler stahlfaserbewehrter Hochleistungsbetone. Beton S Fertigteil-Jahrbuch 2002, pp. 159–166. [301] Nause, P.: Brandverhalten von hochfestem Beton. Beton S Fertigteil-Jahrbuch 2003, pp. 150–151. [302] Schmidt, M., Fehling, E., Bornemann, R. et al.: Ultra-Hochfester Beton: Perspektive fu¨r die Betonfertigteilindustrie. Beton S Fertigteil-Jahrbuch 2003, pp. 10–23. [303] Hegger, J.; Bertram, G.: Spannbetontra¨ger aus ultrahochfestem Beton. Beton S Fertigteil-Jahrbuch 2008, pp. 85ff. [304] Teutsch, M., Steven, G., Empelmann, M.: UHPFRC – ein Baustoff fu¨r MEGA-Druckglieder. Beton S Fertigteil-Jahrbuch 2007, pp. 74–80. [305] Graubner, C.-A., Mu¨ller-Falcke, G., Kleen et al.: Selbstverdichtender Beton fu¨r Fertigteile. Beton S Fertigteil-Jahrbuch 2002, pp. 132–133. [306] Ludwig, H.-M., Hemrich, W., Weise, F., Ehrlich, N.: Selbstverdichtender Beton-Grundlagen und Praxis. Beton S Fertigteil-Jahrbuch 2002, pp. 113–131. [307] Reinhardt, H.-W.: Selbstverdichtender Beton. Beton S Fertigteil-Jahrbuch 2002, pp. 75. [308] Grunert, J. P., Strohbach, C.-P., Teutsch, M.: Vorgespannte stahlfaserversta¨rkte Bauteile aus SVB ohne Betonstahlbewehrung. Beton S Fertigteil-Jahrbuch 2005, pp. 48. [309] Strohbach, C. P.: Langzeitversuch mit stahlfaserbewehrten Spannbetonbindern aus selbstverdichtetem Beton ohne konventionelle Bewehrung. Beton S Fertigteil-Jahrbuch 2008, pp. 102. [310] Neumann-Venevere, P.: Beschleunigte Erha¨rtung des Frischbetons von Tafelelementen durch InfrarotStrahlungsheizung. Betonstein-Zeitung 1971, No. 1, pp. 22–27; No. 2, pp. 74–82. [311] Cement Industry Research Institute. Merkblatt fu¨r die Herstellung geschlossener Betonoberfla¨chen bei einer Wa¨rmebehandlung. beton 1967, No. 4, pp. 142. [312] Wischers, G.: Anmerkungen zum Merkblatt fu¨r die Herstellung geschlossener Betonoberfla¨chen bei einer Wa¨rmebehandlung. beton 1967, pp. 101–103 & pp. 139–142. [313] Wiehrig, H.-J.: Kurzzeitwarmbehandlung von Beton. Betonwerk S Fertigteil-Technik 1975, No. 9, pp. 418–423; No. 10, pp. 492–495. [314] Neubarth, E.: Fru¨hhochfester Beton durch Kurzzeit-Warmbehandlung. Betonstein-Zeitung 1969, No. 9, pp. 536–542. [315] Cement Industry Research Institute. Merkblatt fu¨r die Anwendung des Betonmischens mit Dampfzufu¨hrung. beton 1974, No. 6, pp. 344-346. [316] Neck, U.: Auswirkungen der Wa¨rmebehandlung auf Festigkeit und Dauerhaftigkeit von Beton. beton 1988, No. 12, pp. 488–493. [317] Rieche, G.: Instandsetzung von Stahlbeton bei Scha¨den infolge Korrosion der Bewehrung. Deutsche Bauzeitschrift 1982, pp. 1011–1017. [318] Piguet, A.: Einfluß der Nachbehandlung auf die Betonkapillarita¨t unter Beru¨cksichtigung des Wasserzementwertes und der Festigkeit. International Colloquium for Materials Science & Building Refurbishment, Esslingen, 1983, pp. 149–152.

References

257

[319] Pu¨hringer, P., Wenzlaff, K.: Fertigteilproduktion in Saudi-Arabien mit angeschlossener automatischer Sandstrahlanlage. Betonwerk S Fertigteil-Technik 1980, No. 7, pp. 442–449. [320] van Acker, A.: Neue Oberfla¨chentechniken bei Fertigteilen aus Beton- und Stahlbeton. Beton- und Fertigteiltechnik 1986, No. 9, pp. 556–562. [321] DVS – German Welding Society, Directive DVS 0302, Flammstrahlen von Beton. DVS, 1985. [322] Menzel, U.: Warmbehandlung von Beton. Betonwerk S Fertigteil-Technik 1991, No. 12, pp. 92–97. [323] Janhunen, P.: Neue Herstellungstechnologie fu¨r Fassadenelemente aus Architekturbeton von hoher Qualita¨t. Betonwerk S Fertigteil-Technik 1993, No. 10, pp. 53–66. [324] Fachvereinigung Deutscher Betonfertigteilbau e.V. in BDB: Hinweise zur Erzielung einer ordnungsgema¨ßen Bewehrung von Betonfertigteilen. Betonwerk S Fertigteil-Technik 1985, No. 7, pp. 473–478. [325] Hu¨tten, P., Pasberg, M.: Zweckma¨ßige Ausfu¨hrungsformen von Bu¨gelko¨rben aus geschweißten Betonstahlmatten. Betonwerk S Fertigteil-Technik 1979, No. 10, pp. 633–636. [326] Schwarz, S.: Praktischer Einsatz vorgefertigter Bewehrungsko¨rbe. Betonwerk S Fertigteil-Technik 1980, No. 9, pp. 571–574. [327] National Technical Approval No. Z-1.2-155, BSt 500 WR (B) in Ringen. DIBt. [328] Kulessa, G.: Herstellung und Verarbeitung von Betonstahl in Ringen. Betonwerk S Fertigteil-Technik 1987, No. l, pp. 14–18. [329] Kro¨mer, R.: Rationalisierung im Betonwerk bei der Bearbeitung von Betonstahl vom Ring. Betonwerk S Fertigteil-Technik 1987, No. 1, pp. 23–36. [330] Riechers, H.-J.: Betonstahl in Ringen. Verfahrensweisen und bisherige Entwicklung. Betonwerk S Fertigteil-Technik 1987, No. 1, pp. 19–22. [331] Schwarzkopf, M.: Moderne Bewehrungstechnik. Betonwerk S Fertigteil-Technik 1991, No. 2, pp. 58–60. [332] Ehlert, W., Fuchs, W.: Probleme bei der Direktansteuerung von Biegeautomaten. Betonwerk S Fertigteil-Technik 1991, No. 2, pp. 61–66. [333] Koncz, T.: Neuentwicklungen in der Spannbett-Technik. Betonwerk S Fertigteil-Technik 1981, No. 11, pp. 700–705. [334] Scott, N. L.: The long-line pretensioning method. FIP notes 1985, No. 4, pp. 2–10. [335] Dietl, W.: Fertigteilkonstruktion aus Spannbeton. PORR-Nachrichten 1981, No. 85/86. [336] Schmalhofer, O.: Interessante Spannbetonbauten aus Fertigteilen. Betonwerk S Fertigteil-Technik 1978, No. 4, pp. 198–204. [337] Ruhnau, J., Kupfer, H.: Spaltzug-, Stirnzug- und Schubbewehrung im Eintragungsbereich von Spannbett-Tra¨gern. Beton- und Stahlbetonbau 1977, No. 7, pp. 175–203; No. 8, pp. 204–208. [338] Bruggeling, A. S. G.: bertragen der Vorspannung mittels Verbund. Beton und Stahlbeton 96 (2001) No. 3. [339] Nitsch, A.: Spannbetonfertigteile mit teilweiser Vorspannung aus hochfestem Beton. Chair & Institute for Solid Construction, RWTH Aachen University, Dissertation, 2001. [340] Hegger, J., Nitsch, A.: Neuentwicklungen bei Spannbetonfertigteilen – aktuelle Forschungsergebnisse und Anwendungsbeispiele. Beton S Fertigteil-Jahrbuch 2000, pp. 95–109. [341] Hegger, J., Will, N., Kommer, B. et al.: Einsatz von selbstverdichtendem Beton fu¨r vorgespannte Bauteile. Research report, DAfStb V 416, 2006. [342] Bechert, H.: Vorspannen mit sofortigem Verbund auf der Grundlage der neuen DIN 4227 Teil 1. Betonwerk S Fertigteil-Technik 1980, No. 9, pp. 2–5. [343] Wo¨lfel, E., Kru¨ger, F.: Verbundverankerung von Spannsta¨hlen-Zulassungspru¨fung und Anwendungsbedingungen. Mitteilungen IfBt 1980, No. 6, pp. 162–164. [344] FIP: Report on prestressing steel: anchorage and application of pretensioned 7-wire strands. FIP Report, 1978.

258

5 References

[345] Uijl den, J. A.: Verbundverhalten von Spanndraht-Litzen im Zusammenhang mit Rißbildung im Eintragungsbereich. Betonwerk S Fertigteil-Technik 1985, No. l, pp. 28–36. [346] Bruggeling, A. S. G.: Die bertragungsla¨nge von Spannstahl bei Vorspannung mit sofortigem Verbund. Betonwerk S Fertigteil-Technik 1986, No. 5, pp. 298–302. [347] Lange, H., Paral, J.: Teilweise vorgespannte Spannbetonbinder –Bemessungstabellen in Anlehnung an DIN 1045. Beton- und Stahlbetonbau 1983, No. l, pp. 12–16. [348] Walraven, J. C.: Lastverteilung und Bruchverhalten von Spannbetonhohldecken. Betonwerk S Fertigteil-Technik 1992, No. l, pp. 57–63. [349] FIP: Precast prestressed hollow-core floors. Thomas Telford, London, 1988. [350] Walraven, J. C.: Tragwirkung von Hohlplatten und Hohlplattendecken. Presentation, Stuttgart, 1993. [351] Hosser, D., Richter E., Hollmann, D.: Entwicklung eines vereinfachten Rechenverfahrens zum Nachweis des konstruktiven Brandschutzes bei Stahlbeton-Kragstu¨tzen. Research report, Braunschweig TU, Nov 2008.

Index A Acid-washing treatment 230 Anchor – façade fixings 116 – retaining anchors 117 – suspension anchors 124 B Bearing – category 181, 188 – elastomeric 181 – length 148 – local pressure 182 – pressure 147, 155 Bound stress 221 Box girder 52 Building joint 31 Building services 133 Butt joint 175 C CE marking 9, 11 Column joint 176 Concrete – coated 231 – coloured 225 – fair-face 226 – fibre-reinforced 224 – glass fibre-reinforced 126 – self-compacting 219 – steel fibre reinforced 219 – surfaces 231 – textile-reinforced 127, 219, 225 – ultra-high performance (UHPC) 126, 219, 222 Connections 130 Construction product act 9 Construction products directive (CPD) 9 – attestation of conformity 10 Continuous boot 154 Corbel 69, 95, 96, 122, 169 D Double-headed stud 149 Double-leaf wall 103, 119 Double-T floor unit 62, 132, 155, 215 Dowels 197 Ductility factor 40

E Earthquake, response spectrum 40 Elastomeric bearing 181 Extruder 205 F Façades – anchorages 116 – ceramic tiles 233 – column-type 105 – fenestrate 105 – fixings 124 – horizontal ribbon 105 – panels 126 – U-shaped 105 Factor of safety 43, 159 Fair-face concrete 226 Flame cleaning 231 Floor diaphragm 32, 33, 37, 43, 45, 62, 64, 66, 194, 205 Floor plank 62, 81 – prestressed 85 Floor slab 209 Foundation 99 – pad foundation 163 – pocket foundation 99, 163, 165 Frame 51, 68, 76 H Heat treatment 241 J Joints 31, 63, 112, 174 – amount of reinforcement 206 – longitudinal reinforcement 64 – loop reinforcement 209 – vertical 55 – waterproofing 113 – with a hard bearing 175 – with a soft bearing 175 L Lattice beam 82, 98, 143 – T-beam slab 140 – composite plank floor 172 Local bearing pressure 182


260 M Movement joint 31 Multi-layer separating wall 168 N Natural vibration 42 Notch 150, 193 Notched beam ends 169, 240 O Out-of-plane shear forces 63, 203 P Perimeter tie 64, 72 Plastic bar spacer 169 Prefabricated units – fit calculation 20 – tolerances 16 – – costs 17 – transport 21 Prestressed hollow-core slab 62, 78, 135, 170, 186, 205, 211, 215, 217, 239 R Restraint forces 29 S Sandblasting 230 Sandwich panel 107, 116, 118 – corner detail 118 – thermal insulation 107 Screwed (socket) joint 72 Second-order theory 58, 61, 160 Segmented hollow box 52 Self-compacting concrete 219 Shear connector system 198 Shear dowel 195 Shear friction theory 203 Shear joint 140, 204, 205 Shear wall 29, 33, 49, 54, 69, 194, 204 Skeleton construction 29, 37, 68 Sliding bearings 187 Stability 28, 45 – of building 47 Steel fibre reinforced concrete 219 Stiffening 28, 45 – core 29, 58, 67 – wall 67, 71 Stud 193 Susceptibility to vibration 34 Synthetic fibres 225

Index T Textile-reinforced concrete 127, 219, 225 Thermal expansion coefficient 44 Thermal insulation 103 Torsion 154 – moment 163 Torsional vibrations 39 Torsional resistance 53 Transport fixing 220 U Ultra-high performance concrete (UHPC) 126, 219 Ultra-high strength concrete 126, 219 University of Riyadh 96, 111, 237 V Vapour barrier 103 W Wall – element 98 – double-leaf 103, 119 – multi-layer separating wall 168 – stiffening 67, 71 Warpening stiffness 54 Welded joint 190 Welding methods 190 Z Zu¨blin 6, 83, 219, 225 – 6M system 90, 131, 136, 211 – House 24, 32, 66, 96, 106, 115, 122, 138

Precast Concrete Structures - Hubert Bachmann

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