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Government of Hong Kong
First published, March, 1990
J f,
Prepared by :
•22ZJ:2
Geotechnical Control Office, Civil Engineering Services Department, 6th Floor, Empire Centre, 68, Mody Road, Kowloon, Hong Kong*
This publication is available from : Government Publications Centre, General Post Office Building, Ground Floor, Connaught Place, Hong Kong.
Overseas orders should be placed with : Publications (Sales) Office, Information Services Department, 1, Battery Path, Central, Hong Kong. Price in Hong Kong : HK$30 Price overseas : US$6.50 (including surface postage) Cheques, bank drafts or money orders must be made payable to HONG KONG GOVERNMENT
FOREWORD
This document presents the results of a review carried out within the Geotechnical Control Office of the state of the art in the design of excavation support systems, and in the prediction of ground movements around excavations. Whilst the document is not intended to be a design guide, the information given here should facilitate the use of modern methods and knowledge in the design of excavation support systems. Practitioners are encouraged to comment to the Geotechnical Control Office on the contents of this publication. The review has been carried out under the general direction of Mr J.B. Massey as a project in the Geotechnical Control Office's Research and Development Theme on Excavations. It was undertaken by Mr C.Y. Kwan, with checking, editing and production by Dr P.L.R. Pang. Numerous staff of the Geotechnical Control Office and the Housing Department, and several firms of consulting engineers, have contributed during the preparation of this document. In particular, Dr S. Buttling provided valuable comments in the final stages. The contributions of all those who have assisted are gratefully acknowledged.
i f
A.W. Malone Principal Government Geotechnical Engineer March 1990
CONTENTS Page No. TITLE PAGE
1
FOREWORD
3
CONTENTS
5
1.
INTRODUCTION
9
2.
TYPES OF EXCAVATION SUPPORT SYSTEMS
11
3.
CONSTRUCTION ASPECTS OF EXCAVATION SUPPORT SYSTEMS
13
3.1
SHEET PILE WALLS
13
3.2
SOLDIER PILE WALLS
15
3.3
DIAPHRAGM WALLS
16
3.4
CAISSON WALLS
18
3.5
GROUTED PILE WALLS
19
3.6
STRUTTED OR BRACED WALLS
20
3.7
TIED-BACK OR ANCHORED WALLS
21
4.
5.
6.
LIMIT STATE METHOD OF DESIGN
23
4.1
GENERAL
23
4.2
ULTIMATE LIMIT STATES
23
4.3
SERVICEABILITY LIMIT STATES
24
AN OVERVIEW OF ANALYTICAL METHODS
25
5.1
TYPES OF METHODS
25
5.2
LIMIT ANALYSIS 5.2.1 General 5.2.2 Upper Bound Methods 5.2.3 Lower Bound Methods
25 25 25 26
5.3
ASSOCIATED STRESS AND VELOCITY FIELDS
26
5.4
BEAM ON ELASTIC FOUNDATION 5.4.1 General 5.4.2 Power Series Method 5.4.3 Finite Difference Method 5.4.4 Distribution Method 5.4.5 Discrete Element Method 5.4.6 Advantages and Limitations
5.5
BOUNDARY ELEMENT METHOD
29
5.6
FINITE ELEMENT METHOD
30
CAm"ILEVERED OR UNBRACED WALLS
33
6.1
33
GENERAL
27 ,27 27 27 27 28 28
Page No.
7.
8.
V 9.
6.2
ULTIMATE LIMIT STATES 6.2.1 Types of Limit States 6.2.2 Overall Stability 6.2.3 Foundation Failure 6.2.4 Hydraulic Failure
33 33 34 36 37
6.3
SERVICEABILITY LIMIT STATES
37
6.4
EARTH PRESSURES FOR STRUCTURAL DESIGN
38
6.5
DESIGN WATER PRESSURES
40
STRUTTED OR BRACED WALLS
43
7.1
SINGLE-LEVEL STRUTTED WALLS 7.1.1 General 7.1.2 Ultimate Limit States 7.1.3 Serviceability Limit States 7.1.4 Earth Pressures for Structural Design 7.1.5 Design Water Pressures
43 43 43 44 44 48
7.2
MULTI-LEVEL STRUTTED WALLS 7.2.1 General 7.2.2 Ultimate Limit States 7.2.3 Serviceability Limit States 7.2.4 Earth Pressures for Structural Design 7.2.5 Design Water Pressures
48 48 48 49 50 53
7.3
TEMPERATURE EFFECTS
53
TIED-BACK OR ANCHORED WALLS
55
8.1
SINGLE-LEVEL TIED-BACK WALLS 8.1.1 General 8.1.2 Ultimate Limit States 8.1.3 Serviceability Limit States 8.1.4 Earth Pressures for Structural Design 8.1.5 Design Water Pressures
55 55 55 58 58 59
8.2
MULTI-LEVEL TIED-BACK WALLS 8.2.1 General 8.2.2 Ultimate Limit States 8.2.3 Serviceability Limit States 8.2.4 Earth Pressures for Structural Design 8.2.5 Design Water Pressures
59 59 60 60 61 63
GROUND MOVEMENTS AROUND EXCAVATIONS
j
65
9.1
GENERAL
65
9.2
FACTORS INFLUENCING MOVEMENTS 9.2.1 Effect of Stress Changes 9.2.2 Dimensions of Excavation 9.2.3 Soil Properties 9.2.4 Initial Horizontal Stress / 9.2.5 Groundwater Conditions / 9.2.6 Stiffness of Support System/ 9.2.7 Effect of Preloading / 9.2.8 Effect of Passive Soil Buttresses
65 65 65 65 66 66 66 66 67
Page No.
9.3
9.2.9 Associated Site Preparation Works 9.2.10 Influence of Construction Factors
67 68
PREDICTIVE METHODS 9.3.1 Types of Methods 9.3.2 Empirical Methods 9.3.3 Semi-empirical Methods 9.3.4 Method of Velocity Fields 9.3.5 Finite Element Method
68 68 69 72 72 72
/ /
10.
REVIEW OF LOCAL DESIGN PRACTICE
75
11.
GEOTECHNICAL CONTROL OF EXCAVATIONS IN HONG KONG
77
12.
CONCLUSIONS AND RECOMMENDATIONS
79
REFERENCES
81
TABLES
99
LIST OF TABLES
101
TABLES
103
FIGURES
109
LIST OF FIGURES
111
FIGURES
117
1.
INTRODUCTION
This document presents a review of the state of the art in the design of excavation support systems in soils, and in the prediction of ground movements around excavations. It is written in the context of the dense urban setting of Hong Kong and its geological and topographical characteristics. While a fair proportion of the literature reviewed is based on British and North American work, relevant Hong Kong literature is also cited. A brief review of the design practice and geotechnical control of excavations in Hong Kong is given in Chapters 10 and 11 respectively. In this document, the term "excavation" is applied generally to cover all basements, large pits and trenches, while the term "deep excavation" is reserved for excavations which are deeper than about six~lnetres (20'feet), in conformity with the convenient distinction made by Terzaghi & Peck (1967). Descriptive terms used for rocks and soils are as defined in Geoguide 3 : Guide to Rock and Soil Descriptions (GCO9 1988).
11
TYPES OF EXCAVATION SUPPORT SYSTEMS
Excavations for urban building development and civil engineering works are often carried out up to the property boundary and usually require batters which are not stable without structural support. Some of the structural systems which have been used to provide lateral support to the ground are illustrated in Figure 1 (Institution of Structural Engineers, 1975) 5 and their advantages and disadvantages are listed in Table 1. The following retaining wall types are commonly used in Hong Kong to support excavations : (a) (b) (c) (d) (e)
sheet pile wall, soldier pile wall, diaphragm wall, caisson wall, and grouted pile wall.
The above retaining wall types can be further divided into the following three categories according to the form of support provided : (a) (b) (c)
cantilevered or unbraced wall, strutted or braced wall, and tied-back or anchored wall.
The factors to be considered in the choice of a support system for a deep excavation are given in Table 2. The cantilevered wall depends upon its own strength and embedment to carry the earth and water loads, while strutted and tied-back walls transfer much of these loads to additional support elements. Cantilevered walls are limited to relatively shallow excavations unless designed to be massive and embedded in stiff soils. However, under such circumstances the strutted or tied-back wall is likely to be more economic than the cantilevered wall. In a strutted excavation, there is an economic incentive to the contractor to excavate as far as possible below the supports. For tlerback systems.* the excavation floor at • any given time is a necessary arvEt convenient platform for the installation of the anchors, which is usually carried out at the correct levels due to difficulties in providing access later. In this respect, tie-back systems have an advantage over strutted systems in that they automatically guard against over-excavation. Although an open excavation area is created by the tie-back systems, their adoption is not always possible due to encroachment into adjacent property. In the urban areas of Hong Kong, (the strutted excavation is the most popular system. ")A summary of the design considerations for strutted and tied-back walls is given in Table 3.
13
3.
3A
CONSTRUCTION ASPECTS OF EXCAVATION SUPPORT SYSTEMS
SHEET PILE WALLS
The sheet pile wall is the most common type of temporary retaining wall system used in Hong Kong. When selecting sheet piles to be used, it is important to taken into account the drivability of the piles. The ability of a sheet pile to penetrate the ground depends on the section size of the pile and the type of pile hammer used, as well as the ground conditions, Gillespie (1983) considered that it is difficult to drive sheet piles through soils with' Standard Penetration Test (SPT):':|NI values greater than about 50, and that the penetration through zones with N values of 80 or more is very limited. For heavy pile sections, however, the corresponding SPT values could be up to 80 and 120 respectively. It should be noted that hard driving conditions will necessitate the use of pile sections larger than those required to resist bending moments due to lateral earth pressures. It is not uncommon in Hong Kong for sheet piles not to reach design penetration. At an estimated 75% of sites where sheet piles were used (Malone, 1982), this problem was sufficiently extensive to necessitate a significant design change after the piles were driven. In weathered rocks and colluvium, hard corestones and boulders are the obstacles, whereas in fill, the problems are boulders, old foundations and seawalls. These obstructions are normally removed in advance of the main excavation by means of a local cofferdam, and then the sheet piles are redriven to the required depth of penetration. In some cases, even the use of a powerful pile hammer and heavier steel sheet pile sections is unable to either split or push aside the boulders or corestones. Special measures such as the provision of grout curtains behind, and driving of H-piles in front of the inadequately penetrating sheet piles may be necessary. Sometimes, part of a line of sheet piles is replaced by a group of soldier piles installed in pre-drilled holes through the obstruction. Seepage may be expected to pass through interlocking steel sheet piling which supports a difference in hydraulic head. In compressible soils, the drop in piezometric levels may induce cjyjsol.Tdation. Tearing of interlocks under hard driving conditions may cause ground loss due to groundwater infiltration through the torninterlocks. Extraction of steel sheet piles from cohesive soils often results in removal of soil at the same time, leading to settlement of adjacent ground. Driving steel sheet piles in loose sandy soils can result in settlements in adjacent ground. The effects of noise and vibration caused by sheet piling operations have been discussed by Clough & Chameau (1980) and Sarsby (1982). In Hong Kong, legislation was introduced in 1988 to control noise from construction and other activities. Under the Noise Control Ordinance (Government of Hong Kong, 1988), construction noise permits must now be obtained from the Environmental Protection Department prior to the use of percussion piling during specified periods. Information on 'quiet1 construction equipment and working practices, including percussion piling methods, is given by Environmental Protection Department (1989). Malone (1982) reported six sheet pile wall collapses and two cases of gross movement of sheet pile walls (Figure 2 ) , and listed the factors present in these eight cases as follows :
14 (a)
inadequate toe support to piles which do not penetrate below bottom excavation level (Figure 2(a)), four known instances,
(b)
buckling of struts due to inadequate horizontal or vertical s t i f f e n i n g (Figure 2(b}), three known instances,
(c)
raking struts too steep and improperly fixed to piles (Figure 2(a)), two known instances,
(d)
raking struts inadequately founded (Figure 2(b)), one known instance,
(e)
over-excavation in the passive soil buttress or i t s complete premature removal p r i o r to installation of struts (Figure 2(c)), two known instances.
(f)
In one case, translationai shear failure appears to have taken place along a surface at about toe level in a layer of clay, presumed to be a marine deposit, the low shear strength of which had not been appreciated.
I t should be noted that in seven of these cases lateral support was supposed to be provided by raking struts, and that in at least f i v e of the eight cases the piles involved had not been driven to the design level before excavation commenced. Other case histories of sheet pile wall failures have been given by Sowers & Sowers (1967), Broms & S t i l l e (1976), and Daniel & Olson (1982)• These studies show that failures of anchored bulkheads and excavation supports are seldom the result of the inadequacies of modern earth pressure theories. Instead, they are caused by the neglect of surcharge loads, construction operations that produce excessive earth pressures, poorly designed support systems, inadequate allowance f o r d e f l e c t i o n , deterioration and corrosion, and poor design and construction d e t a i l s . The following comments were given by Beattie (1983) on the practical design of sheet piling under Hong Kong conditions : (a)
Sheet piles are rarely installed in a straight line or v e r t i c a l at depth and the design of shoring and packing may need to be modified on site to cope with the deviations. The packing details should be adequate to take the design loads. Otherwise, yielding of the support w i l l allow movement of the piles and retained ground.
(b)
Guidance should be given on the drawings of actions to be taken i f the sheet piles are prevented from reaching their design depths by b o u l d e r s or o t h e r o b s t r u c t i o n s . Other differences between the designer's assumptions and the actual site conditions should be investigated.
15
3.2
(c)
The designer should specify monitoring, the results of which should be made immediately available if they are to be of use in gauging the performance of the works.
(d)
The site staff should take time to secure progress by ensuring proper installation of struts and support before proceeding with excavation.
(e)
Regular site inspections by the designer and close cooperation with the builder are important to ensure the works comply with the intent of the design at all stages.
SOLDIER PILE WALLS
Soldier pile walls have two basic components, soldier piles (vertical component) and lagging (horizontal component). Soldier piles provide intermittent vertical support and are installed before excavation commences. Due to their relative rigidity compared to the lagging, the piles provide the primary support to the retained soil as a result of the arching effect. Spacing of the piles is chosen to suit the arching ability of the soil and the proximity of any structures sensitive to settlement. A spacing of 2 to 3 m is commonly used in strong soils, where no sensitive structures are present. The spacing is reduced to 1 to 2 m in weaker soils or near sensitive buildings. It is essential that the soldier piles are maintained in full contact with the soil. For this reason, they are either driven or placed in pre-drilled holes, which are backfilled to the ground surface with lean concrete. The lagging serves as a secondary support to the soil face and prevents progressive deterioration of the soil arching between the piles. It is often installed in lifts of 1 to 1.5 m, depending on the soil being supported and on the convenience of working. There are four methods of transferring earth pressures from the lagging to driven soldier piles, as shown in Figure 3 (Peck, 1969b) : (a)
Lagging is wedged against the inside flanges (Figure 3(a)). In this method, the soil yields as it is trimmed to make room for the lagging. Most of the earth pressure is transferred through the soil to the pile and very little through the lagging to the pile. However, any remaining voids between the soil and the lagging could permit further soil movement.
(b)
If large voids exist between the soil and the lagging in (a), they can be filled with grout or mortar to provide full contact (Figure 3(b)).
(c)
Lagging set between spacers, the soldier piles and the soil permits the piles to be included as permanent reinforcement to the structural wall (Figure• 3(c)). Earth pressure is transferred to the piles as the excavation is carried to the face of the piles, but the heavily loaded soil behind
16
the piles is then cut away to provide room for the spacers and the lagging. As a result, the "earth wall 1 moves into the-excavation which can result in substantial surface settlement (Peck, 1969b). Therefore, this method should not be used unless large settlement adjacent to the excavation is acceptable. (d)
Contact sheeting is placed against the flanges of the soldier piles on the wall of the cutting, permitting a tight f i t because the exposed soil can be trimmed cleanly to the edge of the flange (Figure 3(d)J. I t is the most satisfactory method for reducing lateral movements, and also serves to provide a form for concrete for the permanent wall.
I f there is risk of groundwater building up behind a lagged w a l l , or of washing in of soil particles, drainage holes may be provided or small gaps may be left between the boards to allow drainage, and in such cases a f i l t e r material such as a synthetic fabric may be used to prevent loss of soil. Suitable conditions for soldier pile walls are provided by overconsolidated clays, in all soils above the water table i f they have at least some cohesion, and in homogeneous, free-draining soils that can be effectively dewatered. Unsuitable conditions include squeezing soils such as soft clays and running soils such as loose sands. In Hong Kong, s o l d i e r p i l e walls are used mainly f o r small excavations. They are less popular than sheet pile walls, although soldier piles installed in p r e - d r i l l e d holes are often used to overcome the problem of inadequate penetration of sheet piles in areas where rock is encountered at shallow depths below the excavation. They are generally unsuitable for excavations below the groundwater level, particularly where the retained materials are granular in nature. This is because dewatering may cause significant loss of material and unacceptable settlement to adjacent structures and land, and provision of a grout curtain behind and beneath the wall to strengthen the ground and to act as a groundwater cutoff is expensive. For driven soldier piles, driving of the H-sections may cause heave and settlement of nearby structures (D'Appolonia, 1971). In hard driving conditions caused by boulders or other obstructions, soldier piles can be redrilled i f necessary, or relocated to avoid the obstruction, which gives them an advantage over sheet pile walls. Predrilling is often used in urban areas to avoid the noise and vibration of pile d r i v i n g . 3.3 DIAPHRAGM WALLS The diaphragm walling technique was f i r s t introduced to Hong Kong for the construction of the New World Centre (Tamaro, 1981). As a result of i t s successful u t i l i z a t i o n , the technique was used extensively in the construction of the Mass Transit Railway (MTR) underground stations (Openshaw & Marchini, 1980; Mclntosh et a l , 1980; Budge-Reid et a l , 1984) and of deep basements for high-rise buildings (Openshaw, 1981; Chipp, 1983; Craft, 1983; Fitzpatrick & Will ford, 1985; Humpheson et a l , 1986).
17
The design and construction of diaphragm walls have been discussed in detail by Xanthakos (1979) and the practical problems associated with diaphragm walling under Hong Kong conditions have been discussed by Openshaw & Marchini (1980), Davies (1982), and Craft (1983). An important local problem is the temporary stability of the slurryfilled trench. The stability of slurry trenches under Hong Kong conditions has been reviewed by Buttling (1983) and Wong (1984). Wong (1984) considered that both Huder's (1972) method and Schneebeli's. (1964) method are conservative for cases in Hong Kong where the soil is fill, marine deposit or colluvium. The reason for this is possibly due to the local practice of not using the apparent cohesion component of soil strength in design. A design procedure was proposed (Wong, 1984) for predicting the stability of slurry trenches subjected to heavy surcharge load from adjoining and adjacent foundations, by taking into consideration the redistribution of surcharge load to soil adjacent to the excavation, and the redistribution of stresses in the building structure. In reclaimed land in Hong Kong, the presence of randomly dumped loose bouldery fill is not uncommon. A method for dealing with this problem has been described by Craft (1983). The technique is to first excavate a trench (prior to guide-wall construction) along the line of the diaphragm wall by backhoe to below groundwater level, and to backfill it with a lean mix concrete mixed with bentonite. The lean mix design has to be wet enough to be tremied, strong enough to stabilise the boulders, yet weak enough to be excavated subsequently by clamshell, and of low permeability. The guide-walls are then constructed. Excavation by clamshell is carried out through the lean mix concrete and as far as possible into the underlying boulders. Excavation is usually terminated on the basis of declining net progress. At this point, lean mix concrete is tremied into the excavation. This operation is repeated until the bouldery fill stratum has been stabilised by the lean mix concrete. Trench stability in soft clays can be improved by reducing the panel length or increasing the slurry pressure. The slurry p r e s s u r e c a n b e increasedby either increasing the density of the slurry, or raising the level of the slurry above ground level by raised guide-walls; this latter method is not favoured by contractors. To increase the density, special additives can be used, but the properties of the slurry must not conflict with the requirements for placing the concrete. Buttling (1983) considered that viscosity is a more significant parameter than density with regard to displacement of bentonite by concrete, and with proper specification the two parameters can be independently controlled. In sands and other permeable soils, groundwater lowering can be used to improve trench stability, as stability depends on the net slurry pressure, i.e. the difference in head resulting from the difference in level between the slurry within the trench and the external groundwater (Fitzpatrick & Will ford, 1985; Humpheson et al, 1986). During construction of diaphragm walling in weathered granite in Hong Kong, significant ground movements have been observed causing large surface settlement extending a considerable distance from the wall (Davies & Henkel, 1980; Cowland & Thorley, 1984). It is considered that the horizontal movement adjacent to individual panels was a result of swelling of the weathered granite, which created a compressible zone around the diaphragm wall panel. On construction of adjacent panels the arching around the compressible zone broke down, and recompression occurred as the
18
earth pressure built up. This caused horizontal ground movements to extend back from the wall (Davies & Henkel, 1980; Davies, 1982). The ground movements can be controlled by increasing the net slurry pressure supporting the trench, in a similar manner to that described for improving trench stability in marine clays and sands. Corestones or 'bedrock1 (generally taken to be a rock mass containing rock materials wholly or predominantly of grades I to I I I ) encountered during the construction of diaphragm walls in weathered granite can be broken up by grab and chisel (Openshaw & Marchini, 1980). However, diaphragm walling in such cases is extremely expensive due to the cost of rock excavation. Therefore, i t is common practice to stop the wall on top of bedrock. If the toe of the diaphragm wall is exposed as the main excavation continues into bedrock, the wall w i l l need to be underpinned, and chemical grouting may need to be carried out to provide an effective cut-off against groundwater. Diaphragm walling, although expensive when compared with other wall systems, frequently provides the best solution to many underground construction problems, and the technique can usually be adjusted to cope with most conditions of d i f f i c u l t ground and adjacent surcharge by adjustment of the panel length, properties of the slurry, level of slurry in the trench and groundwater lowering. 3.4 CAISSON WALLS Hand-dug caisson retaining walls, often referred to as caisson walls, are widely used in Hong Kong, The caissons are usually dug in stages of about one metre depth, by husband miners with their wives serving as winch operators at the ground surface. Each stage of excavation is lined with a minimum of 75 mm thickness of insitu concrete using a tapered steel shutter suitably supported and designed for ease of striking. The shutter remains in place to provide support for the fresh concrete and surrounding ground while the next stage of excavation proceeds. Submersible electric pumps are commonly used for dewatering at the base of the caissons. Excavation through corestones and other obstructions is carried out by pneumatic drilling. Pope (1983) considered that the main advantages of a caisson wall over a conventional retaining wall are : (a)
It can be constructed without temporary soil cuts or shoring.
(b)
I t requires a small plan area and can be used close to site boundaries and existing structures.
(c)
I t can act as both temporary and permanent retaining structures.
(d) Obstructions, e.g. corestones and boulders, can be overcome without too much d i f f i c u l t y . It should be noted that some of the above advantages also apply to other wall types described in this Chapter. Caisson walls were used in the construction of the MTR underground stations (Benjamin et a l , 1978; Mclntosh et a l , 1980) and of deep basements
19
for multi-storey buildings (Beattie & Mak, 1982; Addington, 1983). The caissons can be contiguous, or closely spaced at about two-diameter centres with in-filled concrete panels. Base heave failures have been experienced with hand-dug caissons in Hong Kong (Chan, 1987). The mechanism of heave is poorly understood, but it is believed to be the result of slaking and swelling, and hence loss of strength, of residual soils and saprolites due to heavy seepage and stress relief. The presence of weak layers such as clayey marine and alluvial deposits may also bring about such failures. The possibility of base heave renders the construction of a deep hand-dug caisson extremely hazardous, particularly when it is carried out below the groundwater table. Moreover, groundwater drawdown for caisson excavation can result in ground settlement and damage to adjacent structures (Morton et al, 1980a & b; 1981). Precautionary measures such as the provision of grout curtains behind the caissons to reduce the effect of drawdown can be expensive. Nevertheless, these are commonly applied whenever unfavourable ground conditions are anticipated. The choice of grouts for hand-dug caisson construction has been discussed by Shirlaw (1987). Another problem with caisson wall construction is the installation of back drainage. Various forms of drainage system have been discussed by Malone (1982) and these are shown in Figure 4. Type I suffers from the defect that at each stage of excavation the work is undermined. In type II, the no-fines concrete is generally not an adequate filter for residual soil/weathered rock. Types III and IV involve placing filter media in a deep, narrow borehole. Type IV also involves the installation of perforated pipes from the caisson shaft as excavation proceeds. Type V requires the installation of conventional horizontal drains. The long-term performance of these drainage systems will depend greatly on the care taken in design, specification, construction, monitoring and maintenance.
3.5
GROUTED PILE WALLS
The grouted pile walling technique was introduced to Hong Kong in the construction of the Tsim Sha Tsui MTR Station (Mclntosh et al, 1980). The most common form that has been used is the "Pakt-in-Place11 or PIP wall system. A PIP pile is formed by boring with a continuous flight hollow shaft auger to the design depth. As the auger is withdrawn, cement-sand mortar is injected through the auger shaft, leaving a formed mortar pile. A reinforcement cage is then lowered into the mortar to form a complete reinforced PIP pile. To form a continuous pile wall, PIP piles (A piles) are formed in alternate pile positions in the above manner. The infill piles (B piles) are then constructed in the same way as the A piles, except that during the auger withdrawal and mortar placement, a high pressure plunger pump is used for injecting cement paste through the pile, which effects a vertical mortar cut-off between the B piles and the adjacent A piles. A recent innovative form of construction, known as '.pipe pile walling1 has been used to overcome the problem of inadequate penetration when driving sheet piles through weathered rock with corestones and through old foundations. A pipe pile is formed by installing a perforated steel casing in a drillhole greater in diameter than the casing. As the drill assembly is withdrawn through the casing, suitable grouts are injected through the perforated casing into the surrounding soil. The forming of a pipe pile
20
w a l l i s s i m i l a r t o t h a t of a PIP w a l l . The advantages of the pipe p i l e w a l l i n g technique are as f o l l o w s :
3.6
(a)
The i n s t a l l a t i o n o f noise and v i b r a t i o n .
the
piles
involves
(b)
The w a l l can p e n e t r a t e t h r o u g h corestones and hard o b s t a c l e s ,
little
boulders,
STRUTTED OR BRACED WALLS
D i f f e r e n t schemes f o r s t r u t t i n g are shown i n F i g u r e 5. These i n c l u d e the c r o s s - l o t s t r u t t i n g and r a k i n g s t r u t support schemes i l l u s t r a t e d i n Figures 5(a) and 5 ( b ) . The diagonal s t r u t t i n g t e c h n i q u e shown i n F i g u r e 5 ( c ) i s more s u i t e d t o s m a l l and p r e f e r a b l y square e x c a v a t i o n s , e . g . s h a f t s - The disadvantage o f c r o s s - l o t s t r u t t i n g i s ^ t h a t t h e working space can be severely r e s t r i c t e d . Diagonal s t r u t t i n g i s sometimes used near t h e corners of wide excavations and serves t o leave a l a r g e p o r t i o n o f t h e excavation open. Because only a few s t r u t s are u s e d , t h e s e s t r u c t u r a l elements can be very l a r g e ( t y p i c a l l y , pipe s e c t i o n s are u s e d ) , and they can be s u b j e c t t o l a r g e v a r i a t i o n s i n l o a d s due t o t e m p e r a t u r e fluctuations. In some cases, t h e s t r u t s are covered or p a i n t e d w i t h a r e f l e c t i v e p a i n t t o minimise temperature e f f e c t s . Various methods have been devised to expedite t h e c o n s t r u c t i o n o f deep basements t h r o u g h t h e use o f t h e permanent s t r u c t u r e t o support t h e excavation during c o n s t r u c t i o n . These methods have t h e advantage o f a f f o r d i n g savings o f both time and r e s o u r c e s , but they do demand a h i g h degree o f c o o r d i n a t i o n between t h e designer and c o n t r a c t o r , and t i g h t c o n t r o l o f t h e sequence o f o p e r a t i o n s . Figures l ( e ) , l ( k ) and l(m) show examples o f such methods. B e a t t i e & Yang (1982) described a p r o j e c t i n which an i n d u s t r i a l b u i l d i n g was constructed using t h e "top down1 method. During c o n s t r u c t i o n , the p a r t i a l l y completed b u i l d i n g was used t o support an e x c a v a t i o n up t o 15m deep. G a r r e t t & Fraser (1984) described how t h e t o p down method ( i n which the columns o f a b u i l d i n g below ground l e v e l are f i r s t c o n s t r u c t e d and t h e n t h e permanent s t r u c t u r e i s b u i l t p r o g r e s s i v e l y downwards, supporting t h e excavation w a l l s as they are exposed), t h e bottom up method ( i n which the excavation is f i r s t completed and t h e permanent s t r u c t u r e i s b u i l t from t h e bottom upwards), and a combination o f t o p down and bottom up techniques, have been used s u c c e s s f u l l y i n t h e c o n s t r u c t i o n o f t h e MTR underground s t a t i o n s . Connections and d e t a i l s are o f c r i t i c a l importance s t r u t t e d e x c a v a t i o n . Shoring plans should show :
in
an
(a)
a p p r o p r i a t e means f o r p o s i t i o n i n g and f i x i n g o f s t r u t s and w a l i n g s , and f o r b r a c i n g o f s t r u t s i n b o t h v e r t i c a l and h o r i z o n t a l planes t o p r o v i d e lateral s t a b i l i t y ,
(b)
d e t a i l s of brackets,
web
and
connection
stiffeners,
and
internally
21
(c)
details of welding and other means of structural connection,
(d)
details of connections between raking struts and the foundation,
(e)
sequence of installation and dismantling elements of the structure, and
(f)
provisions, if any, for wedging and jacking of struts to prevent horizontal movement.
of all
Typical connection details are shown in Figures 6 and 7. Procedures for prestressing, wedging or jacking to maintain tight contact for all support members and to provide for distribution of load to struts and walings are also very important. Typical prestressing details are shown in Figures 8 and 9. Flat jacks are also commonly used in Hong Kong for prestressing strutted excavations. Preloading to about 50% of the design load is common practice in areas where displacements are of concern. Strut removal can give rise to additional displacements, and these may be controlled by a combination of well-planned re-strutting and effective compaction of the backfill between the wall and the structure. For walls supported by raking struts, the Canadian Foundation Engineering Manual (Canadian Geotechnical Society, 1985) recommends a detail which involves the construction of trenches in the passive earth buttresses to accept the raking struts and their footings. It also recommends that the distance of a footing from the wall face should be more than 1.5 times the length of the wall embedded below the toe level of the passive earth buttress. These measures are intended to limit the lateral movement of flexible walls.
3.7
TIED-BACK OR ANCHORED WALLS
Some tie-back or anchored systems are shown in Figure 10. These include 'deadman1 type anchorages, driven piles and grouted anchor systems. In Hong Kong, prestressed ground anchors are commonly used for retaining excavations. The connecting elements between anchorages and anchor heads are either tie rods or cable strands. Because these elements are commonly made of high strength steel, usually only a small area of steel is required to give the necessary capacity; as a result, the longitudinal stiffnesses of the cables or tie rods are relatively small compared to those of struts. Double corrosion protection of these important structural elements is now standard practice in Hong Kong (GCO, 1989). Because of their relatively low cost and ease of construction in confined areas, soil nails are being used increasingly in Hong Kong for supporting excavations (Massey & Pang, 1989).
23
4.
4.1
LIMIT STATE METHOD OF DESIGN
GENERAL
In this document, the approach to geotechnical design is discussed using limit state terminology, A limit state is a performance criterion related to a limiting condition beyond which a structure or an element is assumed to become unfit for its purpose. The limit state approach to design requires the designer to check the adequacy of the structure against possible limit states (e.g. ultimate and serviceability limit states) that could be of importance during the design life of the structure. There is nothing new about the limit state approach; international standards on structural use of concrete based on such an approach were available in the early 1970's. Limit state design is, however, unnecessarily associated with the use of probability theory (in the assessment of risk), statistics of variation (in the assessment of loads and material properties) and partial safety factors. The application of limit state principles to geotechnical design does not demand the use of these, either singly or in combination. Nevertheless, the possible inclusion of these additional considerations in future geotechnical codes of practice has caused a great deal of controversy (see, for example, Boden, 1981; Bolton, 1981; Ovesen, 1981; Semple, 1981; Simpson et al, 1981; Smith, 1981). Limit state design of earth retaining structures is advocated in the Draft Model for Eurocode EC 7 : Common Unified Rules for Geotechnics, Design (European Geotechnical Societies, 1987), which gives a full account of the use of the underlying limit state principles. A structure should be so designed and constructed that its functional requirements are fulfilled throughout its design life. Normally this requirement is satisfied by demonstrating that the structure has the necessary safety against ultimate failure (ultimate limit state) at all times, and that the predicted movements and deformations can be tolerated (serviceability limit state). These two so-called 'functional requirements1 mean that two separate calculations are often performed for each individual structure, viz. an analysis against ultimate states of failure, and a deformation analysis of the states of normal use (working conditions). In practice, however, which of these analyses should be used is often dictated by experience; the other analysis may often either be omitted completely or limited to a rough check.
4.2
ULTIMATE LIMIT STATES
An ultimate limit state of a structure is deemed to have been reached when sufficient parts of the structure, the soil around it, or both have yielded to result in the formation of a failure mechanism in the ground or severe damage (e.g. yielding or rupture) in principal structural components. Design against ultimate limit states includes checking against modes of failure such as sliding, overturning, bearing capacity, uplift, and hydraulic failure. Theoretical methods for analysing structures at collapse limit state have been developed within the framework of plasticity theory. There are two common approaches, the kinematic method proposed by Coulomb and the static method advocated by Rankine. When dealing with perfectly plastic materials, these two approaches provide upper bound and lower bound
24
e s t i m a t e s o f the collapse load r e s p e c t i v e l y . S t a t i c methods, such as stress a n a l y s i s , are known t o be p e s s i m i s t i c when compared w i t h k i n e m a t i c techniques, such as wedge or s l i p c i r c l e a n a l y s i s , which are known t o be inherently o p t i m i s t i c . K i n e m a t i c a l l y a d m i s s i b l e mechanisms are u s e f u l in determining the possible degree o f conservatism i n t h e i r s t a t i c c o u n t e r p a r t s . Where s t a t i c a l l y admissible s t r e s s f i e l d s cannot be e a s i l y d e r i v e d , i t is necessary t o optimise the geometries o f d i f f e r e n t presumed mechanisms to achieve a reasonably low upper bound s o l u t i o n ,
4.3
SERVICEABILITY LIMIT STATES
A s e r v i c e a b i l i t y l i m i t s t a t e o f a s t r u c t u r e is deemed t o have been reached w i t h t h e onset o f e x c e s s i v e d e f o r m a t i o n or o f d e t e r i o r a t i o n . Design a g a i n s t s e r v i c e a b i l i t y l i m i t s t a t e s i n c l u d e s checking a g a i n s t unacceptable t o t a l and d i f f e r e n t i a l movements, c r a c k i n g , and v i b r a t i o n . Serviceability l i m i t s t a t e calculations for earth retaining s t r u c t u r e s i n v o l v e s o l u t i o n o f s o i l - s t r u c t u r e i n t e r a c t i o n problems, which r e q u i r e t h e use o f d e f o r m a t i o n p a r a m e t e r s . F i n i t e element methods i n c o r p o r a t i n g a p p r o p r i a t e s o i l models are w e l l s u i t e d t o t h i s t y p e o f analysis. However, t h e y are p r e s e n t l y t o o cumbersome and c o s t l y f o r g e n e r a l a p p l i c a t i o n , and a c c u r a t e d e t e r m i n a t i o n o f t h e r e q u i r e d s o i l parameters is d i f f i c u l t . C o n v e n t i o n a l ' w o r k i n g s t r e s s 1 design methods o f t e n c o n t a i n l a r g e s a f e t y f a c t o r s against c o l l a p s e which are also intended t o l i m i t t h e e x t e n t o f y i e l d i n g in the s o i l ; e l a s t i c s o i l s t r a i n s are g e n e r a l l y c o n s i d e r e d to be less damaging than p l a s t i c s o i l s t r a i n s . However, t h i s approach does not y i e l d t h e order o f t h e l i k e l y deformation o f t h e s t r u c t u r e and t h e adjacent ground. Design methods a r e c o n s i d e r a b l y c l a r i f i e d i f t h e p r e d i c t i o n o f deformations is kept separate from c a l c u l a t i o n s f o r checking against c o l l a p s e .
25
AN OVERVia OF ANALYTICAL METHODS
5.1
TYPES OF METHODS
Methods f o r a n a l y s i n g e a r t h r e t a i n i n g c l a s s i f i e d i n t o t h e f o l l o w i n g types : (a) (b) (c) (d) (e)
structures
can be
broadly
limit analysis, associated stress and velocity fields, beam on elastic foundation, boundary element method, finite element method.
These methods are discussed in the following sections.
5.2 5.2.1
LIMIT ANALYSIS General
The methods of Coulomb and Rankine are the forerunners of the limit analysis approach. As discussed in Section 4.2, they provide upper bound and lower bound estimates of the collapse load respectively. Therefore, they are well suited to the analysis of ultimate limit states. However, they cannot predict deformations associated with the collapse load nor can they yield information prior to the limit state being reached.
5.2.2
Upper Bound Methods
An upper bound s o l u t i o n is obtained by o p t i m i s i n g the c o l l a p s e load o b t a i n e d from presumed mechanisms w i t h k i n e m a t i c a l l y a d m i s s i b l e v e l o c i t y fields. The v e l o c i t y f i e l d s have t o s a t i s f y t h e v e l o c i t y boundary c o n d i t i o n s , and have t o be continuous except a t c e r t a i n d i s c o n t i n u i t y s u r f a c e s , where t h e normal v e l o c i t y must be continuous but where t h e t a n g e n t i a l v e l o c i t y may undergo a jump on c r o s s i n g t h e boundary. The t r i a l wedge method i s normally used t o c a l c u l a t e e a r t h pressures a c t i n g on an e a r t h r e t a i n i n g s t r u c t u r e (GCO, 1 9 8 2 ) . Janbu (1957) d e v e l o p e d a g e n e r a l i s e d procedure o f s l i c e s method which can be adapted f o r e a r t h pressure computations. Methods o f s l i c e s such as t h a t o f Janbu (1957) w i l l g i v e upper bound s o l u t i o n s f o r v e l o c i t y f i e l d s which are k i n e m a t i c a l l y a d m i s s i b l e . They have advantages i n t h a t l a y e r e d s o i l d e p o s i t s can be a n a l y s e d , d r a i n e d and undrained s o i l s t r e n g t h s can e a s i l y be accommodated, and t h e mechanics o f t h e methods a r e known t o most g e o t e c h n i c a l e n g i n e e r s . A l s o , when programmed i n t o a c o m p u t e r , these methods can g r e a t l y f a c i l i t a t e t h e calculations. The d i s a d v a n t a g e s , which are common t o a l l l i m i t a n a l y s i s p r o c e d u r e s , are : (a)
No i n f o r m a t i o n on deformations can be o b t a i n e d .
(b)
The e f f e c t s o f i n t e r a c t i o n o f m a t e r i a l s w i t h d i s p a r a t e s t r e s s - s t r a i n c h a r a c t e r i s t i c s a r e not considered.
(c)
The i n f l u e n c e o f c o n s t r u c t i o n sequence or unusual i n i t i a l s t r e s s c o n d i t i o n s are not accounted f o r .
(d)
A t r i a l procedure must be employed to l o c a t e t h e most c r i t i c a l f a i l u r e s u r f a c e .
5.2.3 Lower Bound Methods A lower bound solution of the collapse load is determined from a statically admissible stress field which satisfies the stress boundary conditions, is in equilibrium, and nowhere violates the yield condition. If the upper bound and lower bound solutions coincide, then the result is the exact solution for the problem considered. Numerical analysis using lower bound limit procedures was proposed by Sokolovski (1965) following a finite difference scheme. In the Sokolovski approach, the equations of equilibrium in two dimensions are combined with the Mohr-Coulomb failure criterion to give a pair of differential equations for the stresses along stress characteristics (i.e. lines along which the Mohr-Coulomb criterion is satisfied). The resulting differential equations can be solved abyn using finite difference techniques for stresses in the 9 soil stru % !° *?? ; cture interface. The solution yields a ion for the collapse load, the earth pressure distribution on the $ d l S t r i b u t i o n in the soi l mass at the assumed a
* s i s somewhat s i m i l a r t o t h e i s more f l e x i b l e than S o k o l o v s k i ' s t e c h n i q u e , In L v L ^ the s o i l mass adjacent t o the s t r u c t u r e is modelled as ? 2 «PP«-oach e1 f J e n t S k c o n r \ e c t e d a t ™dal p o i n t s , which a r e t h e b o u n d a Srners oTthe St eSS specified at the J y i e l d C r i t^e r icao n d i it ni o nt hs e are f nodes alnnn 5 p < °™.of constraint eouations? L ' c t * m e a n s ° f . l i n e a r Programming t e c h n i q u e s , optimum p lower bound 4 t r e s s e s t w i t h i n t h e elements a r e o b t a i n e d . Because the method Wn reSS f i e l d n would a P P e a r t o be i d e a l l y suited for USP in fs nf af i 1 °$ ?*}f m ° r e C O m p' l e x Problems. However, Lysmer noted tha? the L ^ ] n fc r f ^ v « r y r a p i d l y w i t h t h e ss ii zz ee o o ff the mesh ^ % a p p l l c a t l o n has c u r r e n t l y been l i m i t e d t o
?4
1ower
ana1
proems only
SUffer
the a r e
5.3
bound
more
from d i f
m
^ ^ u l t
o f
t h e
disadvantages o f to apply, t h e i r r S l e in
ASSOCIATED STRESS AND VELOCITY FIELDS
( r $ * •^ development o f c o n s i s t e n t associated) s t r e s s and s t r a i n f r flXed d e m e n t s of structural deflection^r ^ a d To beain a la H° is s P e c i f i ^ , and an i n i t i a l s t r e s s f i e l d is p r e d i c t e d b d ^ T ^ } ° f S o k o l o v s k i (1965) • Next, an i n i t i a l s t r a i n % 1 ^ l d I s as deSCribed b .1 ^ James e t a l (1972) and Serrano ( 1 9 7 2 ^ usino 011 d e f o r m a t i o n A f t e r d e t e r m i n a t i o n o f the s t a i n ? Parameters. assumed s o i l parameters are not
27
strains. An iterative procedure is then adopted to generate stress and strain fields which are consistent with all parametric assumptions. At convergence, a complete distribution of stresses and strains in the soil is obtained, as well as earth pressures on the structure. Fundamental questions remain concerning the appropriateness of the assumed stress-strain laws and the effect on the predictions caused by noncoincidence of principal stress and strain increment directions. Also, questions relating to practical applications, e.g. how wall friction distributions can be assigned, and how account can be taken of construction sequences, seepage loadings, structural flexibility, etc., have yet to be answered. As the method is not fully developed, it is difficult to evaluate its potential as a design tool. Nevertheless, the method does provide the geotechnical engineer with a powerful tool to enhance his understanding of the processes involved in soil-structure interaction.
5.4 5.4.1
BEAM ON ELASTIC FOUNDATION General
The assumption of a beam or slab on an elastic foundation has found application in numerical analysis of sheet piles and strutted excavations. Power series, finite difference, distribution and discrete element methods have been employed for the solution of the governing differential equations. In each case, the elastic foundation is assumed to generate a reactive pressure proportional to the deflection (Winkler f s (1867) hypothesis) and, in essence, consists of a bed of springs. The soil response is usually characterised by a spring constant, which is related to the 'coefficient of subgrade reaction1. The coefficients of horizontal subgrade reaction recommended by Terzaghi (1955) are normally used. The following Sections give a brief review of work based on the Winkler model.
5.4.2
Power Series Method
Rowe (1955a) used t h e power s e r i e s method t o s o l v e t h e f l e x u r e e q u a t i o n f o r c a n t i l e v e r e d and anchored sheet p i l e w a l l s and presented t h e r e s u l t s i n t h e form o f design charts s u i t a b l e f o r design o f f i c e use.
5.4.3
Finite Difference Method
Palmer and Thompson (1948) presented a finite difference method for solving the flexure equation for a sheet pile wall. The solution is of the iterative type and requires a knowledge of the coefficient of subgrade reaction and boundary conditions at the extremities of the sheet pile. An important characteristic of this method is that the coefficient of subgrade reaction need not be a constant but can be a function of depth or pressure.
5.4.4
Distribution Method
Turabi and Balla (1968a) presented a distribution theory of analysis for the solution of the flexure equation based on the approximate modelling of the soil below excavation level by five linear springs. The formulae
28
obtained are in a form suitable for general use, and tables and charts are available for design purposes. 5.4.5 Discrete Element Method Haliburton (1968) presented an efficient discrete element solution for a flexible wall on nonlinear foundation. The wall was divided into short equal lengths of beam elements, each of which could rotate without bending. The elements could transmit transverse forces and couples, and therefore transverse and rotational restraints could be applied. Each element could also sustain an axial force that contributes only to the bending moment but does not produce any axial deformation. The earth pressure-deflection relationship of the soil is assumed to be nonlinear and a function of depth. For simplicity, the portions of the curve between at-rest and passive states and between at-rest and active states were approximated by two straight lines with different slopes (Figure 11). Independent curves for the soil on the active and passive sides of the wall were developed, and the two curves were then superimposed to obtain a combined curve (Figure 12). With the use of a computer, repeated t r i a l and adjustment procedures were carried out until there was no significant difference between two subsequent deflected shapes of the wall* Rauhut (1969), in a discussion of Haliburton's paper, showed that in general Haliburton's method did not give good agreement with experimental results due to the omission of wall friction and the linear distribution of deflection. Bowles (1988) also modelled the sheet pile wall as beam elements, but utilised the concept of subgrade reaction for soil below the excavation level. The matrix stiffness method of analysis was used to obtain the solution. Ng (1984) demonstrated how a program for the matrix method can be implemented on a programmable calculator. 5.4.6 Advantages and Limitations An advantage of the 'beam on elastic foundation1 approach over limit analysis is its ability to account for structure flexibility and soil stiffness. Thus the effects of stress redistribution in soil as a result of differential structural deflections are accommodated. Although the magnitude of the shear forces and bending moments in the wall and the strut or anchor loads are not very sensitive to the values of spring stiffness (constant) used in the analysis, the predicted deformations of the supporting wall are. The limitations of the 'beam on elastic foundation1 approach are : (a) There is inherent difficulty in determining the appropriate spring stiffnesses (constants) of the soil for analysis as these are not fundamental soil properties. (b) The method cannot directly simulate construction sequence, unusual initial .soil stresses, and development of wall friction.
29
(c)
The method does not yield surface movements behind the retaining structure.
Haliburton's method of utilising the relationship between lateral earth pressure and displacement is a slightly better way of modelling soilstructure interaction. However, it should be noted that although the influence of the horizontal earth pressure at rest on the deformation of the wall is taken into account, the uncoupled nature of the soil model and the difficulty in establishing the relationship between lateral soil pressure and deformation limits the applicability of this method in practice.
5.5
BOUNDARY ELEMENT METHOD
Wood (1979) developed a boundary element computer program to analyse pile groups subject to lateral loads and subsequently extended the program to include the analysis of multi-propped sheet pile and diaphragm walls. The wall was assumed to be a series of discrete beam elements attached to the soil at common nodes. Only the compatibility of horizontal displacements between the wall and the soil was maintained at these nodes. The soil was modelled as an inhomogeneous elastic continuum utilising an approximate extension of Mindlin's solution (Mindlin, 1936), based on the assumption that the displacement is a function of the local values of the elastic parameters. The initial disturbing force was determined from a consideration of the coefficient of earth pressure at rest and the out-ofbalance earth pressure on either side of the wall. The earth pressures were derived and compared with the active and passive pressure envelopes obtained from classical theory. An iterative procedure was adopted to ensure that the final soil reaction always lies within the active-passive range. No account was taken of arching. At each excavation stage the incremental movements were computed and the summation of the incremental movement results were given. The application of this program to the prediction of the performance of a diaphragm wall is given by Wood & Perrin (1984). A similar program was also developed by Pappin et al (1985). The soil on each side of the wall was modelled as an elastic solid, the flexibility of which was generated either by way of the integrals of the Mindlin equations, or by interpolation and sealing of flexibility matrices calculated for a simplified soil profile using finite element methods. A semi-empirical formulation was also developed to allow for variations in soil stiffness with depth. The wall was modelled by a series of elastic beam elements. In addition, the earth pressures were limited to within active and passive limits. Arching is permitted by considering the soil forces acting on the wall compared with the forces consistent with possible failure surfaces within the soil. Other features that can be accommodated by the program include struts, anchors and the effects of surcharges. This program has been applied to the prediction of the performance of the basement for the new Hong Kong Bank headquarters building (Fitzpatrick & Will ford, 1985; Humpheson et al, 1986). The boundary element method developed by Pappin et al takes arching into account, and is more rigorous than Wood's approach in the formulation of the soil-structure interaction problem. It also appears to be able to overcome the shortcomings of Haliburton's method. Boundary element programs, although not as general as the finite element methods, have a
30
s i g n i f i c a n t advantage over the l a t t e r i n t h a t they are much cheaper t o apply and s u f f i c i e n t l y accurate f o r most design p r o b l e m s . They are n a r t i c u l a r l y s u i t a b l e f o r parametric s t u d i e s . The shortcoming o f t h e boundary element methods is t h e i r i n a b i l i t y t o p r o v i d e i n f o r m a t i o n on the surface movement behind r e t a i n i n g s t r u c t u r e s . Other methods are r e q u i r e d to c o r r e l a t e the d e f l e c t i o n s of r e t a i n i n g s t r u c t u r e s w i t h s u r f a c e movements (see Section 9 . 3 ) .
5.6
FINITE ELEMENT METHOD
A complete method f o r s o l v i n g s o i l - s t r u c t u r e i n t e r a c t i o n problems is the f i n i t e element method which i n c o r p o r a t e s a continuum s o i l model. This method o f f e r s the designer an a n a l y t i c a l t o o l which can s i m u l a t e almost a l l of the complex facets of the s t r u t t e d or t i e d - b a c k w a l l except u n q u a n t i f i a b l e v a r i a b l e s such as workmanship or g e o l o g i c a l u n c e r t a i n t i e s . The techniques involved in the a p p l i c a t i o n o f t h e f i n i t e element method t o t h e a n a l y s i s o f s o i l - s t r u c t u r e i n t e r a c t i o n have been discussed by Clough (1972a), Clough & Tsui (1977) and t h e I n s t i t u t i o n o f S t r u c t u r a l Engineers (1989). T s u i & Clough (1974) showed t h a t t h e assumption o f plane s t r a i n usually adopted i n f i n i t e element a n a l y s i s o f s t r u t t e d and anchored w a l l s cannot be a r b i t r a r i l y e x t r a p o l a t e d to many p r a c t i c a l p r o b l e m s , p a r t i c u l a r l y those i n v o l v i n g s o l d i e r p i l e w a l l s or l i g h t sheet p i l e w a l l s « Diaphragm w a l l s , however, were shown to approximately s a t i s f y t h e c o n d i t i o n s assumed in a plane s t r a i n a n a l y s i s . For a l l types o f w a l l s t h a t have p r e s t r e s s e d s u p p o r t s , a c h a r t (which is reproduced i n Figure 13) can be used t o p r o v i d e a q u a l i t a t i v e i n d e x t o t h e a p p l i c a b i l i t y o f a plane s t r a i n a n a l y s i s . The d e v i a t i o n of t h r e e - d i m e n s i o n a l pressures from a plane s t r a i n d i s t r i b u t i o n i s defined i n terms o f t h e s o i l s t i f f n e s s , t h e w a l l s t i f f n e s s and the t i e - b a c k spacing. The c h a r t shows t h a t t h e s t i f f e r t h e w a l l , t h e closer the t i e - b a c k spacing, and t h e s o f t e r t h e s o i l , t h e more a c c u r a t e i s the assumption o f plane s t r a i n c o n d i t i o n s . To s i m u l a t e d i s c o n t i n u o u s w a l l elements such as s o l d i e r p i l e s , s t r u t s or t i e - b a c k s i n a plane s t r a i n a n a l y s i s , the bending s t i f f n e s s o f t h e s o l d i e r p i l e s i n t h e w a l l and t h e a x i a l s t i f f n e s s o f the s t r u t s or t i e - b a c k s have t o be represented on a ' u n i t l e n g t h 1 b a s i s . Analyses using t h i s r e p r e s e n t a t i o n y i e l d r e s u l t s t h a t are c h a r a c t e r i s t i c o f the average values between those a t t h e supports and those a t t h e mid-point between t h e s u p p o r t s . Clough and Mana (1976) demonstrated t h a t s i m u l a t i o n o f e x c a v a t i o n i n a f i n i t e element analysis may be s a t i s f a c t o r i l y accomplished by a number o f d i f f e r e n t procedures, provided t h a t a p p r o p r i a t e i n p u t s o i l parameters are used. Both a nonlinear e l a s t i c and an e l a s t o - p l a s t i c s o i l model can be made t o a r r i v e a t t h e same p r e d i c t e d excavation b e h a v i o u r . However, changes i n excavation support d e s i g n , which w i l l b r i n g about u n a n t i c i p a t e d performance, o f t e n take place d u r i n g c o n s t r u c t i o n . I s o l a t i o n of the f i n i t e element analysis from the c o n s t r u c t i o n stages can e a s i l y lead t o erroneous performance p r e d i c t i o n s , because i m p o r t a n t l a t e r occurrences a r e not accounted f o r . Hence, the f i n i t e element a n a l y s i s should be used i n p a r a l e i w i t h an o b s e r v a t i o n a l procedure (Peck, 1969a), and allowance s h o u l d be made f o r r e - a n a l y s e s t o be p e r f o r m e d t o accommodate s o i l parameter re-estimates and c o n s t r u c t i o n changes. Updated s o i l parameters are best obtained by comparing e a r l y observed performance w i t h comparable f i n i t e ^ e l e m e n t predictions. The design engineer should p r e f e r a b l y be involved d u r i n g c o n s t r u c t i o n t o monitor t h e performance o f t h e e x c a v a t i o n ,
31
verify the design assumptions, and to effect any necessary amendments to the design. The two single most important variables required in a finite element analysis are the stiffness of the ground, which mainly affects the displacements, and the magnitude of the initial horizontal stresses, which affect both the displacements and the bending moments. It has been found that the values of the soil stiffness determined from small-scale insitu tests or from laboratory tests are less than those back-analysed from field measurement of the performance of an excavation. The reduction in stiffness has been attributed to sample disturbance or ground disturbance upon insertion of the testing equipment. However, recent field and laboratory studies have demonstrated that conventional methods of strain measurement can lead to very significant underestimates of soil stiffness. Some soils have been found to exhibit nonlinear stressstrain behaviour, with a stiffness at small strain (about 0,01%) higher than that normally measured in conventional laboratory tests (at about 0.1% strain). The influence of this small strain nonlinearity in soil-structure interaction has been studied by Jardine et al (1986) using a finite element analysis. The nonlinear soil model that was used predicted a pronounced settlement trough close to a strutted excavation, which agrees well with that observed in the field. However, such behaviour cannot be predicted by the conventional linear elastic soil model. Martin (1977) showed that the deformation modulus of soils derived from insitu weathering of igneous and metamorphic rocks can be approximated by the pressuremeter modulus, and that a linear relationship exists between the logarithm of Standard Penetration Test (SPT) 'N1 value and the logarithm of pressuremeter modulus. A similar relationship has also been found for Hong Kong residual soils and saprolites. The Young's modulus, E, of such soils is usually correlated with the N values using pressuremeter tests (Chiang & Ho, 1980), plate loading tests (Sweeney & Ho, 1982) and full-scale tests. Endicott (1984) showed that the E/N ratio for completely weathered granite back-analysed from the performance of excavations is greater than that obtained from pressuremeter tests, which in turn is greater than that from triaxial tests on undisturbed samples of completely decomposed granite. Chan and Davies (1984) have given a chart which shows the correlation of soil modulus with SPT JN1 values (Figure 14). Ranges of E/N ratio have also been given by Fraser*(1985), Ku et al (1985), Ku et al (1985) and Whiteside (1986). The insitu horizontal stress in the ground is related to the insitu vertical stress by K o , the coefficient of earth pressure at rest. From conventional soil mechanics, it is well known that K o , being dependent on stress history, is not a fundamental soil property. For a- soil in the normally consolidated (i.e. low K o ) state, the stress change required- to reach the active state is small compared with that needed to mobilise passive r e s i s t a n c e . The contrary is t r u e for a soil in the overconsolidated (i.e. high K o ) state (Figure 1 5 ) . Therefore, the horizontal displacements of the wall and surface settlements behind the wall are dependent on the magnitude of the insitu stress. The existence of a high insitu horizontal stress would likely result in large wall displacements and surface settlements. For excavated walls in soils with a high initial K o value (K o ^2), the prop forces and wall bending moments can greatly exceed those predicted by limit equilibrium methods (Potts & Burland, 1983b; Potts & Fourie, 1984).
32
Geoauide 1 : Guide t o R e t a i n i n g Wall Design (GCO, 1982) recommends t h a t the value o f Ko should be not less than 0.5 f o r t h e d e s i g n o f walls r e t a i n i n g Hong Kong s o i l s d e r i v e d from i n s i t u rock w e a t h e r i n g This is desDite a mean Kn value o f 0.32 f o r ' u n d i s t u r b e d ' c o m p l e t e l y decomposed q r a n i t e obtained by Chan (1976) i n t h e l a b o r a t o r y . I t c o u l d be argued t h a t such low values were due t o t h e i r r e v e r s i b l e e f f e c t s o f s t r e s s r e l i e f . However t h e r e i s other evidence which suggests t h a t Ko c o u l d be t h i s low i n r e s i d u a l s o i l s and s a p r o l i t e s . From t h e observed performance of e x i s t i n g s t r u c t u r e s , Lumb (1979) s u g g e s t e d t h a t a c t i v e and a t - r e s t pressures i n Hong Kong r e s i d u a l s o i l s and s a p r o l i t e s s h o u l d be s m a l l e r than would be estimated by conventional c a l c u l a t i o n s . E n d i c o t t (1982) obtained earth pressure c o e f f i c i e n t s o f 0.19 and 0.17 f o r w a l l movements o f 6 mm and 12 mm r e s p e c t i v e l y i n a completely weathered g r a n i t e and deduced an a t - r e s t p r e s s u r e c o e f f i c i e n t between 0.21 and 0 . 2 5 . The weakening t h e o r y f o r r e s i d u a l s o i l s and s a p r o l i t e s proposed by Vaughan & Kwan (1984) has predicted t h a t Ko should be lower than t h a t obtained from J a k y ' s c l a s s i c a l r e l a t i o n s h i p Ko = 1 - s i n 0 ' . Howat (1985) measured very low minimum values of a c t i v e earth pressure c o e f f i c i e n t Ka i n g r a n i t i c s o i l s and showed t h a t Ko pressures may be estimated by t h e f o l l o w i n g f o r m u l a (Howat & C a t e r , 1985), which takes i n t o account t h e unconfined compressive s t r e n g t h , q u : (7h - u w = ((7V - u w - q u ) ( l where (Th = h o r i z o n t a l t o t a l a\\ = v e r t i c a l t o t a l
- s i n 01)
(1)
stress,
stress,
u w = pore water pressure. The above equation supports Vaughan & Kwan's p r e d i c t i o n less than (1 - s i n 0 1 ) .
that
Ko should be
Although t h e data a v a i l a b l e from t h e l i t e r a t u r e suggest t h a t t h e a t rest earth pressure o f Hong Kong r e s i d u a l s o i l s and s a p r o l i t e s c o u l d be low, no d i r e c t measurements o f Ko i n t h e f i e l d have been r e p o r t e d . Ko values may be deduced by back-analysing c a r e f u l l y i n s t r u m e n t e d e x c a v a t i o n s , but such values are l i k e l y t o be on t h e low s i d e because, by n a t u r e o f excavations, movements can t a k e p l a c e even when \/ery s t i f f s u p p o r t s are provided. The use o f s e l f - b o r i n g pressuremeters f o r measuring t h e i n s i t u h o r i z o n t a l s t r e s s should be c o n s i d e r e d . However, i t s h o u l d be noted t h a t t h e i n s e r t i o n o f an instrument i n t o s a p r o l i t e s may d e s t r o y any bonding and a l t e r t h e i n s i t u s t r e s s e s . T h e r e f o r e , t h e e f f e c t o f d i s t u r b a n c e s h o u l d be assessed. The advantage o f the f i n i t e element method i n t h e a n a l y s i s o f e a r t h r e t a i n i n g s t r u c t u r e s l i e s i n i t s a b i l i t y t o p r e d i c t both e a r t h pressures and deformations w i t h a minimum o f s i m p l i f y i n g assumptions. Both s t r u c t u r e a d s " ° V . . a r e considered i n t e r a c t i v e l y , so t h a t t h e e f f e c t s o f s t r u c t u r a l f l e x i b i l i t y a r e t a k e n i n t o account. L i m i t a t i o n s i n u s i n g t h e method p r i m a r i l y d e r i v e from t h e i n a b i l i t y t o p r e s c r i b e a p p r o p r i a t e c o n s t i t u t i v e laws f o r t h e s o i l , and t o d e t e r m i n e t h e p a r a m e t e r s needed f o r t h e c o n s t i t u t i v e models. However, when u t i l i s i n g t h e o b s e r v a t i o n a l procedure JSfiw . , « * ? ? ' w • f l . n l t e element method o f a n a l y s i s can be expected t o S i 1 c t r , t + „ *lgn 1 + n f ° n n a t T a n d accurate performance p r e d i c t i o n s f o r s o i l - s t r u c t u r e i n t e r a c t i o n problems.
33
6.
6.1
CANTILEVERED OR UNBRACED WALLS
GENERAL
A cantilevered wall derives its support from the passive resistance below the excavation level (or 'dredge line1) in order to balance the active pressure from the soil above that level. This type of wall is economical only for moderate wall heights, since the required section modulus increases rapidly with increase in wall height (the bending moment increases with the cube of the cantilevered height of the wall}* The lateral deflection of this type of wall can be relatively large. Lowering of the excavation level, e.g. by construction activities, erosion and scour in front of the wall, has to be carefully controlled, since the stability of the wall depends primarily on the passive resistance developed in front of the wall. Cantilevered walls of the flexible type, such as sheet pile walls, are normally used as temporary supports, e.g. support for the initial stage of a strutted excavation. When used as permanent supports, they are normally of the rigid type; in Hong Kong, these usually take the form of caisson walls (see Section 3.4). When used to retain an excavation of substantial height, unpropped caisson walls are usually socketed into rock. In the design of cantilevered walls, there are four primary areas of geotechnical consideration : (a) (b) (c) (d)
ultimate limit states, serviceability limit states, earth pressures for structural design, and design water pressures.
In many cases, the consideration of serviceability limit states governs the geotechnical design.
6.2 6.2.1
ULTIMATE LIMIT STATES Types of Limit States
The following ultimate geotechnical design :
limit
states
are
usually
considered
(a)
Overall instability : the embedment depth of a cantilevered wall should be adequate to prevent overturning of the wall.
(b)
Foundation failure : the cantilevered wall should penetrate a sufficient depth to prevent base heave in cohesive soils.
(c)
Hydraulic failure : the depth of penetration should be adequate to prevent hydraulic failure (i.e. piping or heave) in cohesionless soils.
in
34
6.2.2
Overall S t a b i l i t y
The o v e r a l l s t a b i l i t y o f c a n t i l e v e r e d w a l l s is u s u a l l y assessed using l i m i t e q u i l i b r i u m methods o f analysis in which t h e c o n d i t i o n s o f f a i l u r e are p o s t u l a t e d , and a f a c t o r o f s a f e t y a p p l i e d t o ensure t h a t such a f a i l u r e s t a t e does not occur. A c a n t i l e v e r e d wall cannot be i n e q u i l i b r i u m w i t h o u t i t s lower end being f i x e d , as shown in Figure 1 6 ( a ) . For a c a n t i l e v e r e d w a l l r o t a t i n g about a p o i n t near i t s t o e , the t h e o r e t i c a l pressure d i s t r i b u t i o n is o f t h e f i x e d - e a r t h form, as shown in Figure 1 6 ( b ) , when t h e w a l l is a t t h e p o i n t of c o l l a p s e . The pressure d i s t r i b u t i o n shown is c o n s i d e r a b l y i d e a l i s e d , p a r t i c u l a r l y at the p o i n t o f r o t a t i o n , 0 , at which i t i s assumed t h a t t h e r e is an instantaneous change from f u l l passive pressure i n f r o n t o f t h e w a l l to f u l l passive pressure behind the w a l l . The c a l c u l a t i o n o f t h e depth o f embedment corresponding to t h i s pressure d i s t r i b u t i o n i n v o l v e s e q u a t i n g t h e h o r i z o n t a l forces and t a k i n g moments about 0 t o o b t a i n two e q u a t i o n s w i t h two unknown depths, d and z (Figure 16(b)) • The e q u a t i o n s o b t a i n e d are g e n e r a l l y r a t h e r c o m p l i c a t e d , one being a q u a d r a t i c and one a cubic expression i n both d and z . The s o l u t i o n s f o r d and z are u s u a l l y o b t a i n e d by a process o f i t e r a t i o n . I n view o f the considerable a l g e b r a i c complexity o f t h e f u l l method, the s i m p l i f i c a t i o n i l l u s t r a t e d i n Figure 16(c) i s w i d e l y used i n p r a c t i c e . This s i m p l i f i c a t i o n assumes t h a t t h e d i f f e r e n c e between t h e p a s s i v e resistance at t h e back o f the w a l l and t h e a c t i v e p r e s s u r e i n f r o n t acts as a concentrated f o r c e , R, a t t h e t o e . By t a k i n g moments about t h e t o e (thereby e l i m i n a t i n g the f o r c e R from t h e e q u a t i o n ) , t h e depth o f embedment in the s i m p l i f i e d model, d 0 > is found. Because o f t h i s s i m p l i f i c a t i o n , t h e value o f d 0 is s l i g h t l y less than the value o f d o b t a i n e d from t h e f u l l method and i s more l i k e l y t o be nearer t o d - z / 2 . To account f o r t h i s , t h e usual p r a c t i c e i s t o increase d 0 by a small amount (up t o 20%) • A s i m p l e check is u s u a l l y made t o ensure t h a t t h e a d d i t i o n a l embedment is s u f f i c i e n t to provide a f o r c e at l e a s t as l a r g e as t h e assumed f o r c e , R. T h i s can be achieved from a simple c o n s i d e r a t i o n o f h o r i z o n t a l e q u i l i b r i u m . I n most c a s e s , a 20% i n c r e a s e in embedment w i l l be c o n s e r v a t i v e . This s m a l l a d d i t i o n a l increase i n the value o f d 0 is s p e c i f i c a l l y t o account f o r t h e s i m p l i f i c a t i o n , r a t h e r than to provide a f a c t o r o f s a f e t y . There are a number o f ways o f a p p l y i n g t h e f a c t o r o f s a f e t y f o r t h e o v e r a l l s t a b i l i t y o f c a n t i l e v e r e d as w e l l as propped c a n t i l e v e r e d w a l l s embedded in s o i l s . These are : (a)
Factor on embedment d e p t h . A f a c t o r is a p p l i e d to t h e c a l c u l a t e d embedment d e p t h a t limiting e q u i l i b r i u m , as explained i n t h e US S t e e l Sheet P i l i n g Design Manual ( U n i t e d S t a t e s Steel C o r p o r a t i o n , 1975) and t h e BSC General S t e e l P i l i n g Handbook ( B r i t i s h Steel C o r p o r a t i o n , 1986).
(b)
Factor on moments based on gross p r e s s u r e . This m e t h o d i s d e s c r i b e d by t h e I n s t i t u t i o n of S t r u c t u r a l Engineers (1951) and NAVFAC (1982b), and i s g i v e n i n Geoguide 1 (GCO, 1982) for checking the adequacy o f p e n e t r a t i o n o f a s t r u t t e d excavation. I t consists o f f a c t o r i n g only the gross passive pressure. The water pressure is not
35
factored, (c)
F a c t o r on moments based on net t o t a l p r e s s u r e . This method is advocated by t h e BSC General Steel P i l i n g Handbook. The n e t h o r i z o n t a l pressure d i s t r i b u t i o n a c t i n g on the w a l l i s d e r i v e d , and t h e f a c t o r o f s a f e t y is d e f i n e d as t h e r a t i o o f t h e moment o f t h e net passive forces t o t h e moment o f t h e net a c t i v e f o r c e s *
(d)
Factor on moments based on net a v a i l a b l e passive resistance. This method was developed by Bur land et a l ( 1 9 8 1 ) , and is analogous t o t h e c a l c u l a t i o n o f the bearing capacity for a s t r i p load. The f a c t o r o f s a f e t y is taken as t h e r a t i o o f the moment o f t h e net a v a i l a b l e passive r e s i s t a n c e t o t h e moment a c t i v a t e d by t h e r e t a i n e d m a t e r i a l , i n c l u d i n g water and surcharge.
(e)
F a c t o r on shear s t r e n g t h on b o t h a c t i v e and passive s i d e s . The shear s t r e n g t h s o f t h e s o i l s are reduced by d i v i d i n g c 1 and t a n 0 1 by a f a c t o r of safety. These reduced (or m o b i l i s e d ) shear s t r e n g t h parameters are then used t o c a l c u l a t e t h e a c t i v e and passive p r e s s u r e s .
(f)
F a c t o r on shear s t r e n g t h on p a s s i v e s i d e o n l y . This method is recommended by CIRIA ( 1 9 7 4 ) , and is g i v e n i n Geoguide 1 (GCO, 1982) f o r design o f cantilevered walls. The shear s t r e n g t h o f t h e s o i l on t h e p a s s i v e s i d e i s f a c t o r e d , b u t no f a c t o r o f s a f e t y is a p p l i e d on t h e a c t i v e s i d e .
A comparison o f t h e f a c t o r s o f s a f e t y d e f i n e d by methods ( b ) , ( c ) , (d) and (e) was c a r r i e d out by Potts & Burland (1983b) and Symons ( 1 9 8 3 ) . A l s o , t h e f a c t o r s o f s a f e t y d e f i n e d by methods ( a ) , ( b ) , (d) and (e) were compared by P a d f i e l d & Mair (1984). These comparisons r e v e a l t h a t t h e r e is no u n i q u e r e l a t i o n s h i p between t h e r e s u l t s o b t a i n e d by d i f f e r e n t d e f i n i t i o n s of factor of safety. The choice o f method is l a r g e l y one o f c o n v e n i e n c e , and t h e f a c t o r o f s a f e t y i s r e l a t e d t o t h e method used, p r o v i d e d t h a t t h e methods are a p p l i e d c o n s i s t e n t l y . Factors o f s a f e t y a p p r o p r i a t e t o methods ( a ) , ( b ) , (d) and (e) are recommended by P a d f i e l d & Mair ( 1 9 8 4 ) , and these are shown i n Table 4 . The f o l l o w i n g are s p e c i f i c comments r e l a t e d t o each method o f d e f i n i n g factor of safety. Method (a) i s e m p i r i c a l and should always be checked a g a i n s t one o f t h e o t h e r m e t h o d s . Method (b) may g i v e e x c e s s i v e p e n e t r a t i o n s a t lower angles o f shearing r e s i s t a n c e ( 0 ' < 2 0 ° ) . Therefore, d i f f e r e n t f a c t o r s o f s a f e t y are recommended f o r d i f f e r e n t ranges o f 0 1 . Method (c) g i v e s very high f a c t o r s o f s a f e t y compared w i t h o t h e r methods, and i s not recommended f o r design unless an e x c e p t i o n a l l y h i g h f a c t o r o f s a f e t y is s p e c i f i e d » Compared w i t h o t h e r methods, method (d) appears t o g i v e a c o n s i s t e n t lumped f a c t o r o f s a f e t y t h r o u g h o u t t h e p r a c t i c a l range o f s o i l s and w a l l g e o m e t r i e s . U n l i k e method ( d ) , t h e same conceptual framework f o r s l o p e s t a b i l i t y a n a l y s i s can be a p p l i e d i f method (e) i s adopted. However, an anomalous s i t u a t i o n may occur i n methods (e) and ( f ) where s l o p i n g ground o c c u r s . Owing t o t h e d i f f e r e n t f a c t o r s o f s a f e t y
36
adopted f o r slope s t a b i l i t y and f o r o v e r a l l s t a b i l i t y o f t h e w a l l , the angle o f shearing resistance o f t h e s o i l , a f t e r f a c t o r i n g , may be less than the angle of the sloping ground. This can g i v e r i s e t o d i f f i c u l t i e s i n the evaluation of a c t i v e and passive p r e s s u r e s . For method ( f ) , Geoguide 1 (GCO 1982) recommends t h a t a s a f e t y f a c t o r o f 3 s h o u l d be a p p l i e d t o the shear s t r e n g t h o f s o i l on the passive s i d e . This l a r g e f a c t o r o f s a f e t y on shear s t r e n g t h , which is g e n e r a l l y considered t o be c o n s e r v a t i v e , i s not related t o the s t a b i l i t y c r i t e r i o n . I n s t e a d , i t i s a p p l i e d f o r t h e purpose of l i m i t i n g w a l l deformation. Malone (1982) showed t h a t , f o r t h e case o f l e v e l ground on both sides of a cantilevered w a l l w i t h zero water p r e s s u r e , t h e s a f e t y f a c t o r on shear strength (0'=35°) on the passive s i d e can be doubled from 1.5 t o 3.0 by a moderate increase i n s o i l embedment r a t i o (= 1 + d / H , where d and H are as defined i n Figure 16(a)) from 1.8 t o 2 . 1 5 . Pope (1983) showed f o r the same s i t u a t i o n t h a t a s a f e t y f a c t o r o f 1.5 a p p l i e d t o shear s t r e n g t h (0'=4O°) on both a c t i v e and passive sides (method ( e ) ) r e s u l t s i n s m a l l e r embedment depth than by a p p l y i n g a s a f e t y f a c t o r o f 3 on t h e p a s s i v e side only (method ( f ) ) . The same conclusion can be drawn when t h e d e s i g n water l e v e l is at one t h i r d the r e t a i n e d h e i g h t o f t h e w a l l . Bica & Clayton (1989) compared a range o f l i m i t e q u i l i b r i u m design methods f o r c a n t i l e v e r e d w a l l s i n g r a n u l a r m a t e r i a l s . I t was found t h a t a p a r t f r o m t h e d e f i n i t i o n and m a g n i t u d e o f f a c t o r o f s a f e t y , t h e assumptions made i n t h e l i m i t e q u i l i b r i u m a n a l y s i s ( e . g . w a l l f r i c t i o n and method o f c a l c u l a t i n g earth pressures and t h e i r d i s t r i b u t i o n s ) and the c h o i c e o f method f o r d e t e r m i n i n g s o i l and s o i l / w a l l shear s t r e n g t h parameters ( e . g . use o f plane s t r a i n t e s t or t r i a x i a l compression t e s t ) can s i g n i f i c a n t l y a f f e c t t h e r e s u l t s o f d e s i g n . For c a n t i l e v e r e d w a l l s socketed i n t o r o c k , o v e r a l l s t a b i l i t y depends on t h e l a t e r a l load capacity o f t h e rock s o c k e t . The methods used t o assess the load capacity of t h e rock socket were reviewed by Greenway e t al (1986). F a i l u r e o f the rock socket by movement a l o n g d i s c o n t i n u i t i e s i n rock should be considered, as w e l l as bearing f a i l u r e o f t h e i n t a c t rock mass. The persistence of a rock j o i n t i s u s u a l l y yery d i f f i c u l t t o assess during ground i n v e s t i g a t i o n . T h e r e f o r e , i t i s usual p r a c t i c e t o c a r r y out rock j o i n t d i s c o n t i n u i t y surveys d u r i n g c o n s t r u c t i o n .
6.2.3
Foundation Failure
Bottom heave is a problem which primarily occurs in soft to medium clays. The failure is analogous to a bearing capacity failure, the difference being that stresses in the ground are relieved instead of increased. Two types of analysis are available for calculating the factor of ^Sn s t b a s e heave, asa shown in Figure 17. The methdd of Terzaghi fOr 1 ow a n d w i d e e \ Af (195 n£1? ?5 2 , ^avations, whereas that by •1S e x a c *6)i n•1S Ahu i t a b l e f o r deeP and n a r r o w excavations. Neither KJoS. th l . • * *t the effect of any wall segment penetrating ^ S ^ 0 " S a ? tye r ae Qb ayi . rSet s ibsatsi en 9h esaov ei l i movement) is not accounted s usua1 y Q heave i s u s u a 1^ not'les? t h K f ^ V p iH u notles? t h K f ^ V p - i H v a eus h ^ specified s p e c i f i e d to t o be be v a e s h sst trreeffll hh ' l * ? f / 9 t corrected in a d Bierrum u g ™ u' n * l * e d ? n. ft h/ e a n a l 9s ti s c o r r e c t e d i n accordance w iith th stmnoth strength
i f U U ?h ^ l . * ^ u n c o r r e c t e d vane shear i s used, the actual f a c t o r o f s a f e t y w i t h respect t o base heave
37
may be very close to unity (Aas, 1985). Where the safety factor falls below 1.5, substantial wall movement can occur as a result of yielding in the subsoils (Clough et al, 1979; Mana & Clough, 1981). Where limiting movement is the governing criterion, a safety factor not less than 2 is usually required (Clough et al, 1979; Mana & Clough, 1981). If soft clay extends to a considerable depth below the excavation, the beneficial effects of even relatively stiff walls in reducing deformation have been found to be small. However, if the lower portion of the wall is driven into a hard layer, the effectiveness of the wall in reducing deformation is increased appreciably (see Section 9.2.6). A base heave analysis for wide excavation in anisotropic clay was given by Clough & Hansen (1981) and is illustrated in Figure 18. It appears that unless the 'passive strength1 (undrained shear strength at 90° stress reorientation) of a clay is less than 80% of its 'active strength1 (undrained shear strength at 0° stress reorientation), the base heave factor of safety will be within 10% of the value obtained by assuming isotropy with respect to the 'active strength1. The marine soils of Hong Kong appear to show only a slight amount of anisotropy (Lumb, 1977). Hence, anisotropy is normally not considered. The use of field vane shear strength, corrected in accordance with Bjerrum (1973), in an isotropic analysis is probably adequate in most cases. Although it is rare, heave may also occur in cohesionless soils. The factor of safety against base heave in these materials can be estimated from Figure 19.
6.2.4
Hydraulic Failure
Where groundwater exists above the base of the excavation, and if the toe of the cantilevered wall does not penetrate into an impermeable layer, flow will occur under the wall and upward through the base of the excavation. Instability or gross deformation of the base, involving piping or heave, occurs if the vertical seepage exit gradient at the base of the excavation is equal to about unity. To prevent hydraulic failure, the wall must penetrate a sufficient depth below the base of the excavation. Design charts for wall penetration required for various safety factors against heave or piping in isotropic sands and in layered subsoils are given in NAVFAC (1982a). These charts are reproduced in Figures 20 and 21. A design chart is also given by the Canadian Foundation Engineering Manual (Canadian Geotechnical Society, 1985) which shows that the seepage exit gradient for a circular excavation, in the middle section of the sides of a square excavation and in the corners of a square excavation, is 1.3, 1.3 and 1.7 times that for a strip excavation respectively. For clean sand, exit gradients between 0.5 and 0.75 will cause instability of the base. To avoid this, wall penetration for a safety factor of 1.5 to 2 against piping or heave is usually provided.
6.3
SERVICEABILITY LIMIT STATES
The design of cantilevered walls is currently based on limit equilibrium calculations incorporating a high factor of safety to account for both failure and deformation considerations. This approach does not necessarily achieve the aim of limiting deformation (see Section 4.3).
38
The "beam on e l a s t i c f o u n d a t i o n 1 approach i s u s u a l l y employed in deformation analyses. Bowles (1988) modelled t h e sheet p i l e w a l l as beam elements and the s o i l as e l a s t i c s p r i n g s ( t h e s p r i n g s can be n o n l i n e a r ) . Pope (1983) proposed a model f o r deformation analyses i n which t h e ground p r e s s u r e / d i s p l a c e m e n t c u r v e i s approximated by a s e r i e s o f n o n l i n e a r springs and the w a l l as a s e r i e s o f l i n e a r e l a s t i c beam e l e m e n t s . He showed t h a t , by c a r e f u l s e l e c t i o n o f t h e c o e f f i c i e n t o f e a r t h pressure at r e s t and t h e s p r i n g s t i f f n e s s e s ( c o n s t a n t s ) , good agreement w i t h the observed d e f l e c t e d shape could be o b t a i n e d . Pope a l s o showed t h a t t h e r e may be a s u b s t a n t i a l component o f r o t a t i o n f o r a c a i s s o n w a l l t h a t is not socketed i n t o r o c k . He considered t h a t f o r a design based on s t a b i l i t y considerations a l o n e , even by a p p l y i n g a f a c t o r o f s a f e t y o f 3 on s o i l strength on the passive s i d e , unacceptably l a r g e d e f o r m a t i o n s may s t i l l occur i f t h e caissons are not socketed i n t o r o c k . The boundary element method (see Section 5.5) and t h e f i n i t e element method (see Section 5.6) can a l s o be used i n t h e d e f o r m a t i o n a n a l y s e s . These methods are superior t o those using t h e subgrade r e a c t i o n approach, because s o i l - s t r u c t u r e i n t e r a c t i o n e f f e c t s can be b e t t e r taken i n t o account. The l a t e r a l displacements o f the w a l l p r e d i c t e d by l i n e a r e l a s t i c spring models can only be regarded as rough e s t i m a t e s , and need t o be checked by f i e l d measurements.
6.4
EARTH PRESSURES FOR STRUCTURAL DESIGN
I n t h e s t r u c t u r a l design o f t h e c a n t i l e v e r e d w a l l , which i n v o l v e s c a l c u l a t i o n o f bending moments and shear f o r c e s , t h e magnitude as w e l l as the d i s t r i b u t i o n o f earth pressures corresponding t o s t r u c t u r a l f a i l u r e are required. For t h i s purpose, c l a s s i c a l a c t i v e and p a s s i v e e a r t h pressure d i s t r i b u t i o n s , which correspond to t h e l i m i t s t a t e c o n d i t i o n , are u s u a l l y m o d i f i e d i n some a r b i t r a r y manner ( e . g . by t h e d e s i g n s a f e t y f a c t o r against o v e r a l l s t a b i l i t y ) . T h e r e a r e t h r e e a l t e r n a t i v e methods o f c a l c u l a t i n g t h e design bending moment : (a)
The maximum b e n d i n g moment i n t h e w a l l is c a l c u l a t e d at l i m i t i n g equilibrium using unfactored s o i l parameters ( F i g u r e 2 2 ( a ) ) . The depth o f embedment i n t h e design c o n f i g u r a t i o n is greater than t h a t corresponding t o t h e l i m i t i n g equilibrium condition with unfactored parameters, b u t t h e a d d i t i o n a l d e p t h i s n o t used i n t h e calculations. The bending moment d i s t r i b u t i o n a t l i m i t i n g e q u i l i b r i u m is then asssumed t o be equal t o t h a t a c t i n g i n t h e ' w o r k i n g c o n d i t i o n 1 , at which a f a c t o r o f s a f e t y i s provided against s t r u c t u r a l f a i l u r e (Figure 2 2 ( b ) ) .
(b)
An earth pressure d i s t r i b u t i o n , which i s u s u a l l y obtained by a p p l y i n g a s a f e t y f a c t o r t o t h e e a r t h pressures at l i m i t i n g e q u i l i b r i u m , i s assumed t o act on t h e design c o n f i g u r a t i o n o f t h e w a l l . The w a l l b e n d i n g moment d i s t r i b u t i o n is then c a l c u l a t e d by c o n s i d e r i n g moment and f o r c e e q u i l i b r i u m . . Passive pressure at e x c a v a t i o n l e v e l i s , however, not f u l l y m o b i l i s e d i n t h e analyses and hence the r e a l pressure d i s t r i b u t i o n is not
39
properly modelled. The different methods of applying factor of safety to obtain an earth pressure distribution are discussed in Section 6,2.2. (c)
The earth pressure distribution which acts on the design configuration of the wall is calculated by using analytical methods such as those based on the 'beam on elastic foundation' approach (see Section 5.4), and the boundary element and finite element methods (see Sections 5.5 and 5.6).
Methods (a) and (b) are relatively simple for manual computation. However, these methods do not simulate the real soil-wall system, unlike the methods in (c) which take into account the soil-structure interaction. Another advantage of the methods in (c) is that the lateral displacements of the wall required in the displacement analyses are also calculated. If wall deflection is not the prime consideration, method (a) is preferable to method (b) because the passive resistance in front of the wall close to excavation level is likely to be fully-mobilised even under working conditions, and the maximum bending moment is strongly influenced by this (Padfield & Mair, 1984). When method (a) is applied, the design depth of embedment which exceeds that required to give limiting equilibrium is neglected in the calculation of bending moment, as shown in Figure 22 (Padfield & Mair, 1984). For a reinforced concrete cantilevered wall, the reinforcement should not be curtailed at the point where the calculated bending moment is zero. It should be taken to the bottom of the cantilever, and on both faces, to allow for the small reverse bending moments which may occur near the toe. Another problem with the application of method (a) concerns the assumption on water pressures. Since the design embedment depth is greater than that required to give limiting equilibrium, the seepage path round an impermeable wall with the smaller depth of embedment assumed for the bending moment calculation differs from the seepage path corresponding to working conditions. Strictly speaking, the entire embedment depth of the wall should be used in the calculation of seepage pressures, even for the bending moment calculation. However, this tends to be somewhat cumbersome and is usually not allowed for. In general, the error in neglecting the additional seepage length for the calculation of water pressures is rather small. Considerations relating to design water pressures are given in Section 6.5. Bica & Clayton (1989) found that, if plane strain 01 values are used for granular materials, many limit equilibrium methods can give unsafe predictions of maximum bending moments when compared with experimental observations. If the more conservative triaxial compression 0 1 values are used, then a few methods can give reasonable estimates of the moments. It is of interest to note that the available test results show an increase in maximum bending moment as the depth of embedment is increased beyond that required purely for equilibrium. In the structural design of caisson walls, because of difficulty in establishing the 'true1 rockhead levels at design stage, calculations of
40
bending moments and shear f o r c e s are o f t e n c a r r i e d o u t f o r a range of depths t o rockhead. D i f f e r e n t r e i n f o r c e m e n t d e t a i l s a r e shown on the drawings f o r d i f f e r e n t s o c k e t l e n g t h s , t h e d i m e n s i o n s o f which are d e t e r m i n e d by i n s p e c t i o n o f t h e m a t e r i a l s e x p o s e d d u r i n g caisson excavation.
6.5
DESIGN WATER PRESSURES
Proper e v a l u a t i o n o f water pressure and i t s e f f e c t s importance i n t h e design o f excavation s u p p o r t s y s t e m s .
is
of
utmost
For a r e t a i n i n g w a l l which i s impermeable, e i t h e r by n a t u r e o f the w a l l m a t e r i a l or as a r e s u l t o f t h e p r o v i s i o n o f a groundwater c u t - o f f by means o f g r o u t i n g , a d i f f e r e n c e i n water l e v e l may e x i s t behind and i n f r o n t o f the w a l l . Where a r e l a t i v e l y impermeable s o i l l a y e r ( e . g . a marine clay stratum) or an unweathered rock mass o f low p e r m e a b i l i t y is present immediately below t h e t o e o f t h e r e t a i n i n g w a l l , a h y d r o s t a t i c water pressure may act on the w a l l . Groundwater f l o w underneath t h e w a l l can g i v e r i s e t o d i f f e r e n t loading c o n d i t i o n s . The groundwater flow p a t t e r n around an e x c a v a t i o n , which can a f f e c t the water pressures, t h e e a r t h pressures on t h e a c t i v e and p a s s i v e s i d e o f the r e t a i n i n g w a l l , and t h e p i p i n g and heave p o t e n t i a l o f t h e e x c a v a t i o n , depends on the ground c o n d i t i o n s . Figure 23(a) shows a t y p i c a l f l o w - n e t around a r e t a i n i n g w a l l i n homogeneous s o i l under s t e a d y - s t a t e seepage condition. The groundwater t a b l e shown i n t h i s f i g u r e i s c l o s e t o t h e ground s u r f a c e , a s i t u a t i o n which i s commonly encountered i n e x c a v a t i o n s c a r r i e d o u t i n t h e reclaimed areas o f Hong Kong. A s i m p l i f i e d water pressure d i s t r i b u t i o n , as shown i n Figure 2 3 ( b ) , was proposed by P a d f i e l d & Mair (1984) f o r t h e d e s i g n o f c a n t i l e v e r e d and propped w a l l s . This s i m p l i f i e d method assumes t h e h y d r a u l i c head v a r i e s l i n e a r l y down t h e back and up the f r o n t o f the w a l l . This g r e a t l y f a c i l i t a t e s t h e c o m p u t a t i o n o f both the a c t i v e and passive pressures a c t i n g on t h e w a l l . P o t t s & Bur land (1983a) found t h a t t h e r e s u l t s o f a s o i l - s t r u c t u r e i n t e r a c t i o n a n a l y s i s based on such a s i m p l i f i e d method were s i m i l a r t o those from an a n a l y s i s i n which the water pressures had been d e r i v e d from a f l o w - n e t . Kaiser & Hewitt (1982), i n t h e i r d i s c u s s i o n o f t h e e f f e c t o f groundwater f l o w on the s t a b i l i t y and design o f e x c a v a t i o n s , i l l u s t r a t e d t h e i n f l u e n c e o f s o i l l a y e r i n g , h y d r a u l i c a n i s o t r o p y and continuous and discontinuous lenses o f low p e r m e a b i l i t y m a t e r i a l on t h e groundwater f l o w pattern. The r e s u l t a n t water pressures i n these cases can exceed t h a t i n the homogenous s o i l c o n d i t i o n ( F i g u r e 2 4 ) . A knowledge o f t h e s o i l f a b r i c «hLffn+i?+er ° ?, f u t m o s t im P<> r tance by P a d f i e l d & Mair ( 1 9 8 4 ) , who e s S l 1 t « ? £ ^ t V T ! " ° r s a n d P a r t i n 3 s w i t h i n a c l a y s t r a t u m may convey water a t h y d r o s t a t i c pressure t o t h e base o f the w a l l . i h w 2 T n k d e s a \ 1 b e d a b o v e demonstrates t h e need f o r proper ground n S u S . ™ order t o g a i n a good understanding o f t h e ground and o f excavatio t a t h a t V a bclZ* *" th%6QHgn "*For c o n d i t i o n other 9 US th& s i m shown i n Fiautp T ^ h f ° *V\* P l i f i e d w a t e r pressure d i s t r i b u t i o n n e t a n a l v s i / . . . i i ? m l K*. b e , f u f f 1 c 1 e n t l y a c c u r a t e , and a proper f l o w S r o u r S I t e V f l « r ° n r J b l l S h e d i t e c h n 1 q u e s d e s c r i b e d m t r e a t i s e s on and oassivp 2 l U 9 > C e d e r 3 r e 1 n » « 7 7 ) . should be c a r r i e d o u t . The a c t i v e methoTs methods
o T t l P t ^ i r e S H S h ° U l d b e e v a l u a t e d u s i n 9 1im1'* equilibrium ( e . g . t h e t r i a l wedge method and t h e g e n e r a l i s e d procedure o f
41
slices (Janbu, 1957)), with due account taken of the piezometric pressures on the potential failure surface and the soil/wall interface. It should be noted that the presence of water pressures (which reduce effective stresses in soil) will lead to an increase and decrease in active and passive earth pressures respectively. Another adverse effect is the possibility of hydraulic failure (i.e. piping or heave) due to upward hydraulic gradients in front of the retaining wall (see Section 6.2.4). For a retaining wall composed of vertical structural elements spaced horizontally apart (e.g. caissons, soldier piles and pipe piles), some throughflow of groundwater can occur. If the vertical elements are closely spaced, 'damming1 of the groundwater can result in a rise in groundwater level. A method for estimating the effect of piles and caissons on groundwater flow was given by Pope & Ho (1982). Drainage provisions behind an impermeable wall can also affect the flow pattern in the ground (Padfield & Mair, 1984) and their effects should be properly assessed. Infiltration into the ground can sometimes cause a rise in groundwater level. This can increase the water load acting on the excavation support system in some situations. An example of this is where the rise in water level occurs within the retained height, which may take place if the original groundwater level is close to the ground surface or where an impermeable layer is present within the retained soil mass. Even for a retaining wall which has a vertical drainage layer at its back, steadystate seepage can still occur when there is sufficient water infiltrating into the retained soil, causing an increase in pore water pressures and a reduction in soil shear strength. The steady-state seepage condition is most likely to occur in a soil with a relatively low permeability, as rainfall of moderate intensity can supply sufficient water to sustain the seepage flow. The effect of infiltration should be taken into account in design. Deformation of the ground adjacent to excavations may cause breakage of water-carrying services. In some cases, the amount of water released can adversely affect the stability of excavations. Therefore, it is important that the location, size and condition of such services in the vicinity of a proposed excavation are ascertained. In situations where significant water flow can occur in the event of breakage, it will be prudent to allow for the possible increase in water load in the design. Alternatively, consideration should be given to diverting water-carrying services prior to excavation.
43
7.
7.1 7.1.1
STRUTTED OR BRACED HALLS
SINGLE-LEVEL STRUTTED MALLS General
The behaviour of single-level strutted walls is similar to that of single-level anchored walls in which the anchors are installed horizontally. The effect of inclined anchors will be discussed in Section 8.1.1. The factors which influence the behaviour of an anchored flexible wall were reviewed by Bjerrum et al (1972). As a result of wall deflection, the earth pressures acting on a flexible wall such as a sheet pile wall are very much different from those predicted by classical earth pressure theories. On the back of the wall, the pressures at the top and bottom are increased due to the smaller outward deflections at these locations, while the pressures between the excavation and anchor levels tend to be reduced due to the relatively large outward deflections there. This redistribution of earth pressure around a flexible wall may be considered the result of arching. Rowe (1952) found in his model tests on sheet pile walls that the arching effects can be considered to consist of three parts : that due to flexure below the excavation level, that due to flexure above the anchor level, and that due to reduction in earth pressure between the excavation and anchor levels. The flexure below the excavation level is the most stable and reliable source of moment reduction. The flexure above the anchor level could be reduced or destroyed by additional anchor yielding after excavation, and thus cannot be relied upon in all cases. The moment reduction due to pressure reduction between the anchor and excavation levels is unstable and could be destroyed by anchor yield (or elastic shortening of the struts) and backfill settlement, except in the case of piles encastre at the anchorage. As in the design of cantilevered walls, there are four primary areas of geotechnical consideration : (a) (b) (c) (d)
ultimate limit states, serviceability limit states, earth pressures for structural design, and design water pressures.
The c o n s i d e r a t i o n o f u l t i m a t e l i m i t s t a t e s l i m i t s t a t e s u s u a l l y governs t h e d e s i g n .
7.1.2
rather
than
serviceability
U l t i m a t e Limit States
(1) Overall S t a b i l i t y . The o v e r a l l s t a b i l i t y o f s t r u t t e d w a l l s , as i n c a n t i l e v e r e d w a l l s , i s a l s o assessed using l i m i t e q u i l i b r i u m methods o f analysis. T h e r e i s a c h o i c e between adopting a f r e e - e a r t h o r a f i x e d - e a r t h procedure f o r analyzing overall s t a b i l i t y . For a p r o p p e d w a l l , c o n s i d e r a t i o n o f o v e r a l l s t a b i l i t y i n terms o f r o t a t i o n about t h e prop p o s i t i o n i s o n l y a p p l i c a b l e t o f r e e - e a r t h s u p p o r t c o n d i t i o n s . Where f i x e d e a r t h s u p p o r t c o n d i t i o n s a p p l y , provided the w a l l is adequately propped and designed t o r e s i s t t h e shear f o r c e s and bending moments a c t i n g upon i t , t h e r e i s no f a i l u r e mechanism r e l e v a n t t o an o v e r a l l s t a b i l i t y check
44
( P a d f i e l d & M a i r , 1984). Propped w a l l s designed on t h e assumption of f i x e d - e a r t h conditions are t h e r e f o r e not considered h e r e . Assuming f r e e earth support c o n d i t i o n s , the depth o f embedment i s u s u a l l y determined by taking moments about the p o s i t i o n o f t h e prop ( F i g u r e 2 5 ) . The d i f f e r e n t methods o f applying f a c t o r o f s a f e t y are discussed i n d e t a i l i n Section 6.2.2. (2) Foundation Heave and H y d r a u l i c F a i l u r e . C o n s i d e r a t i o n s of foundation heave and hydraulic f a i l u r e are discussed i n S e c t i o n s 6 . 2 . 3 and 6.2.4 r e s p e c t i v e l y .
7*1.3
Serviceability Limit States
As i n the case o f c a n t i l e v e r e d w a l l s , t h e d e s i g n o f propped w a l l s is o f t e n based on l i m i t e q u i l i b r i u m c a l c u l a t i o n s , u s i n g a h i g h f a c t o r of s a f e t y t o ensure adequate s t a b i l i t y and t o r e s t r i c t s o i l and w a l l movements t o acceptable l e v e l s . A more d e t a i l e d approach i s t o employ the 'beam on e l a s t i c foundation 1 approach (see S e c t i o n 5.4) t o e s t i m a t e the l a t e r a l displacements o f t h e w a l l . However, n e i t h e r method takes i n t o account the i n s i t u stress s t a t e o f t h e s o i l p r i o r t o e x c a v a t i o n . Both the boundary element and f i n i t e element methods (see S e c t i o n s 5.5 and 5.6) can be used t o i n v e s t i g a t e t h e i n f l u e n c e o f i n i t i a l s t r e s s e s i n t h e s o i l on the behaviour o f s i n g l e - l e v e l propped r e t a i n i n g w a l l s . An e l a s t o - p l a s t i c c o n s t i t u t i v e law f o r t h e s o i l was employed by Potts & Fourie (1984) i n a f i n i t e element a n a l y s i s t o model t h e behaviour o f a propped w a l l . The r e s u l t s o f t h e i r numerical a n a l y s i s showed t h a t , i f a s u i t a b l e f a c t o r o f s a f e t y a g a i n s t o v e r a l l s t a b i l i t y i s adopted f o r an excavated w a l l i n s o i l w i t h a low Ko value ( K o = 0 . 5 ) , t h e l a t e r a l w a l l movements are o f small magnitude and t h e r e f o r e c o n s i s t e n t w i t h c u r r e n t design philosophy. However, f o r an excavated w a l l i n s o i l w i t h a high Ko value (K o =2), r e l a t i v e l y l a r g e l a t e r a l w a l l movements were p r e d i c t e d , even at shallow excavation depths w i t h l a r g e f a c t o r s o f s a f e t y a g a i n s t o v e r a l l s t a b i l i t y . T h e r e f o r e , i t is i m p o r t a n t t h a t t h e i n s i t u h o r i z o n t a l s t r e s s o f t h e s o i l be c a r e f u l l y assessed i n t h e design o f e x c a v a t i o n s . Where n e c e s s a r y , d e f o r m a t i o n a n a l y s e s s h o u l d be p e r f o r m e d t o e v a l u a t e t h e s e n s i t i v i t y o f t h e design t o t h e assumed value o f K o .
7.1*4
Earth Pressures for S t r u c t u r a l Design
The s t r u c t u r a l d e s i g n o f a propped c a n t i l e v e r w a l l i n v o l v e s t h e c a l c u l a t i o n o f bending moments and shear f o r c e s f o r t h e w a l l and t h e prop f o r c e from t h e d i s t r i b u t i o n o f e a r t h pressures a c t i n g on t h e w a l l . ••" "t 1 ) Wall Bending Moment. There are t h r e e c a l c u l a t i n g t h e design bending moment : (a)
alternative
methods
Assuming f r e e - e a r t h support c o n d i t i o n s , t h e b e n d i n g moment d i s t r i b u t i o n i s c a l c u l a t e d a t limiting equilibrium with unfactored soil parameters, as shown i n Figure 2 6 ( a ) . As i n t h e case of c a n t i l e v e r e d w a l l s , t h e design depth o f embedment i s g r e a t e r than t h a t c o r r e s p o n d i n g t o the l i m i t i n g equilibrium condition with unfactored s o i l parameters, but t h e a d d i t i o n a l depth i s not
of
45
used in the calculations. (b)
An earth pressure distribution, which is usually obtained by applying a safety factor to the earth pressures at limiting equilibrium, is assumed to act on the design configuration of the wall (Figure 26(b)). This is in effect the 'fixedearth support1 analysis. The different methods for applying the factor of safety are discussed in Section 6.2.2. It should be noted that the fixedearth support analysis, which was developed for flexible anchored sheet pile walls embedded in sand (Terzaghi, 1953), may be unsafe when applied to more rigid walls in clays due to the long-term deformation characteristics of such soils.
(c)
The earth pressure distribution which acts on the design configuration of the wall is calculated by using analytical methods such as those based on the 'beam on elastic foundation1 approach (see Section 5.4), and the boundary element and finite element methods (see Sections 5.5 and 5.6).
Rowe (1955a) used a power series solution to solve the flexure equation for sheet pile walls (see Section 5.4.2). On the active side of the wall, triangular and constant active earth pressure distributions are assumed above and below the excavation level respectively. On the passive side, a coefficient of subgrade reaction which increases linearly with depth is assumed. Curves showing the relationship between the degree of flexibility of a sheet pile wall and the reduction of bending moment, expressed as a ratio of the maximum moment calculated assuming free-earth support conditions, were established for both medium dense and very dense granular soils. Rowe's moment reduction theory was recommended in CIRIA (1974). Rowe then extended his flexibility method of analysis to the solution of sheet pile walls encastre at the anchorage (Rowe, 1955b), sheet pile walls at failure (Rowe, 1956), sheet pile walls in clay (Rowe, 1957a) and sheet pile walls subject to a line resistance above the anchorage (Rowe, 1957b). Rowe's work (1952-1957) was reviewed by Lee & Moore (1968). A similar method of analysis was also given by Richart (1955). Schroeder & Roumillac (1983) found that for flexible anchored walls with a sloping excavation line the calculated moments may be reduced by extrapolating Rowe's moment reduction curve. A finite difference method of solving the flexure equation for sheet pile walls was given by Palmer and Thompson (1948). Turabi & Balla (1968a) published a distribution theory of analysis by approximating the soil below the excavation level by five linear springs. A triangular active earth pressure distribution above the excavation level was assumed, similar to that assumed by Rowe (1955a). However, in contrast to Rowe (1955a), the application of a constant active pressure below the excavation level was considered not justified, as the pressure distribution below the excavation level can be obtained from the spring reactions. They later extended their theory by removing the assumption of linear pressure distribution above the excavation level (Turabi & Balla, 1968b). In this respect, their method is an improvement on Rowe's method.
46
Haliburton (1968) presented a d i s c r e t e element method i n which the n o n l i n e a r earth p r e s s u r e - d e f l e c t i o n r e l a t i o n s h i p i s approximated by two straight lines with different slopes. The i m p o r t a n t i n f l u e n c e o f t h e a t rest earth pressure on t h e deformation o f t h e w a l l i s t a k e n i n t o account. Popescu & Ionescu (1977) modified H a l i b u r t o n ' s method by a p p r o x i m a t i n g the earth p r e s s u r e - d e f l e c t i o n r e l a t i o n s h i p by two h y p e r b o l i c c u r v e s . However, Rauhut (1969) commented t h a t , i n g e n e r a l , t h e bending moments c a l c u l a t e d by H a l i b u r t o n ' s method were l a r g e r than those measured i n Rowe's model tests due t o t h e f a c t t h a t w a l l f r i c t i o n and t h e l i n e a r d i s t r i b u t i o n of d e f l e c t i o n s were ignored. A d i s c r e t e element method based on t h e 'beam on e l a s t i c f o u n d a t i o n 1 approach was also proposed by Bowles ( 1 9 8 8 ) . U n l i k e H a l i b u r t o n ' s method, the concept o f subgrade r e a c t i o n was used f o r t h e p a s s i v e r e s i s t a n c e only and, s i m i l a r t o Rowe's method, Coulomb's t h e o r y was used t o c a l c u l a t e the e a r t h p r e s s u r e above t h e e x c a v a t i o n l e v e l . However, c o n s t a n t a c t i v e p r e s s u r e below t h e excavation l e v e l was considered unnecessary, as in Turabi & B a l l a ' s method. Browzin (1982) commented t h a t Bowles' h y b r i d approach i s not l e g i t i m a t e as i t combines c o n d i t i o n s a t f a i l u r e (Coulomb theory) w i t h conditions at working load ('beam on e l a s t i c f o u n d a t i o n ' ) . Maffei et a l (1977b) proposed a model f o r e a r t h r e t a i n i n g s t r u c t u r e s which simulates the step by step c o n s t r u c t i o n o f an e x c a v a t i o n . This model takes i n t o account the e l a s t o - p l a s t i c behaviour o f t h e s o i l as w e l l as i t s h y s t e r e s i s . W i n k l e r ' s hypothesis ( W i n k l e r , 1867) i s adopted i n c a l c u l a t i n g the displacements. Although t h i s model i s a b e t t e r r e p r e s e n t a t i o n o f t h e s o i l - s t r u c t u r e system than t h e o t h e r 'beam on e l a s t i c f o u n d a t i o n 1 methods, the uncoupled nature o f t h e s o i l s p r i n g s cannot model any a r c h i n g e f f e c t o r i g i n a t i n g from other parts o f t h e w a l l . Both t h e boundary element and f i n i t e element methods t r e a t t h e s o i l as a continuum and give t h e best r e p r e s e n t a t i o n o f t h e s o i l - s t r u c t u r e i n t e r a c t i o n . The a p p l i c a t i o n o f t h e boundary element and f i n i t e element methods t o p r e d i c t t h e performance o f propped w a l l s was c a r r i e d out by Wood & Perrin (1984) and by Bjerrum e t a l ( 1 9 7 2 ) , Egger ( 1 9 7 2 ) , Hubbard e t a l (1984), Potts & Burland (1983a) and Smith & Boorman ( 1 9 7 4 ) , r e s p e c t i v e l y . Potts & Fourie (1984) used a f i n i t e element method t o i n v e s t i g a t e t h e influence o f type o f c o n s t r u c t i o n and o f i n i t i a l s t r e s s on t h e b e h a v i o u r o f propped w a l l s (see also Section 7 . 1 . 3 ) . The r e s u l t s o f t h e i r i n v e s t i g a t i o n indicated t h a t , f o r w a l l s i n a low Ko (K o =0.5) s o i l , t h e s i m p l e l i m i t e q u i l i b r i u m c a l c u l a t i o n s produce c o n s e r v a t i v e values o f prop f o r c e and bending moments. For excavated w a l l s i n high Ko (K o =2) s o i l s , prop f o r c e and b e n d i n g moments g r e a t l y exceed t h o s e c a l c u l a t e d u s i n g t h e l i m i t e q u i l i b r i u m approach c u r r e n t l y i n use. I n c r e a s i n g t h e embedment o f t h e excavated w a l l s i n high Ko s o i l s does not reduce t h e prop f o r c e or bending moments, even though t h e f a c t o r o f s a f e t y a g a i n s t o v e r a l l s t a b i l i t y is increased. Large zones o f f a i l e d s o i l , e s p e c i a l l y i n f r o n t o f t h e w a l l , were p r e d i c t e d f o r excavated w a l l s i n high Ko s o i l s , and t h e l a t e r a l e a r t h pressures behind the w a l l d i f f e r s u b s t a n t i a l l y from t h e c l a s s i c a l a c t i v e distribution. Passive c o n d i t i o n s i n f r o n t o f t h e w a l l are c o m p l e t e l y m o b i l i s e d a t s m a l l excavation depths and b e f o r e a c t i v e c o n d i t i o n s a r e approached down t h e back o f t h e w a l l . I n c o n t r a s t , w a l l s i n low Kn s o i l s showed l a t e r a l p r e s s u r e s w h i c h were i n agreement w i t h t h e c l a s s i c a l d i s t r i b u t i o n s . A c t i v e pressures down t h e back o f t h e w a l l are m o b i l i s e d a t small excavation depths and b e f o r e passive pressures f u l l y develop i n f r o n t ot the w a l l . In a l l t h e cases c o n s i d e r e d , high m o b i l i s e d angles o f w a l l
47
friction were predicted. There is therefore a need to assess the K o value in the case of excavated walls. Smith & Boorman (1974) showed that Rowe's model test results can be reproduced using the finite element method. Potts & Fourie (1985) carried out a finite element analysis to investigate the effect of wall flexibility on the behaviour of propped walls. The results showed that, for flexible walls of the sheet pile type embedded in low K o soils, the bending moments are much lower than those given by the simple limit equilibrium predictions. As the wall stiffness increases, the bending moments approach the limit equilibrium value. These results are in full agreement with the model tests of Rowe (1952). However, for stiff walls in high K o soils, the bending moments greatly exceed the limit equilibrium values. A similar trend is observed for the prop force. Curves showing the variation in the ratios of the maximum bending moment and prop force to those predicted by limit equilibrium method with wall stiffness, for various insitu K o values, are given in Figure 27. For excavated walls in high K o soils, it appears that the boundary element and finite element methods are the only suitable design methods. For walls in low K o soils, methods in (c) are also better than methods (a) and (b) because the lateral displacements of the wall as well as the wall bending moments and prop force can be obtained. As deformation is not usually the controlling criterion for propped walls, methods (a) and (b), which are conservative and suitable for hand computation, are normally adopted. In such cases, method (a) is preferable to method (b) because the earth pressure in front of the wall under working conditions is the fullymobilised passive pressure, as revealed by the finite element method of analysis. When method (a) is used, the considerations regarding curtailment of the reinforcement and allowance for water pressure for cantilevered walls apply also to propped walls (see Section 6.4). (2) Prop Force. There are two conflicting requirements concerning the prop positions, viz. the need to install supports at an early stage of excavation to restrict movement and the need to install them at a low enough level to limit the moment acting in the wall (Padfield & Mair, 1984). For greatest economy in the wall section, often props are placed so that moments are equal in the span below and in the cantilevered section above the prop. Shear forces in the wall near the position of the prop may be critical for concrete walls, A factor of safety is usually applied to the design prop force. If method (a) is used for the design moment calculation (see Section 7.1.4(1)), resolution of forces allows a solution to be obtained for the horizontal prop load, which is generally increased by 25% to allow for the possibility of arching and stress redistribution behind the wall (Padfield & Mair, 1984), If method (b) is used for the design moment calculation, the value of the prop load obtained is very sensitive to the factor of safety used. The prop force should always be conservatively assessed because, unlike in the structural design of the wall, the consequences of failure are usually very serious. The system of props and waling should be able to withstand the loss of at least one prop. Raking struts generally allow considerable movement, so it is acceptable to design for active pressure behind the wall, even with several levels of raking struts.
48
7-1.5
Design Water Pressures
Considerations r e l a t i n g are given i n Section 6.5.
7.2
to t h e e v a l u a t i o n
of
design
water
pressures
MULTI-LEVEL STRUTTED HALLS
7,2.1
General
The f a c t o r s which i n f l u e n c e t h e behaviour o f m u l t i - l e v e l flexible s t r u t t e d excavation were reviewed by Bjerrum et a l (1972)* The loads which must be r e s i s t e d by the s t r u t s i n a s t r u t t e d e x c a v a t i o n depend on two entirely different factors. The f i r s t exerted by the s o i l movements shear s t r e n g t h
f a c t o r i s t h e magnitude o f t h e t o t a l a c t i v e e a r t h pressure s o i l behind and beneath the w a l l , down t o t h e d e p t h where are no longer s i g n i f i c a n t . This depends e x c l u s i v e l y on the and the u n i t w e i g h t o f t h e s o i l behind t h e w a l l .
The second f a c t o r i s t h e d i s t r i b u t i o n o f t h e e a r t h p r e s s u r e , which determines how much o f t h e t o t a l a c t i v e e a r t h p r e s s u r e w i l l be c a r r i e d by the s t r u t s . The d i s t r i b u t i o n depends on t h e amount o f a r c h i n g and is c o n t r o l l e d by t h e magnitude o f t h e deformations i n t h e s o i l beneath the excavation r e l a t i v e t o those o f t h e s t r u t s . These d e f o r m a t i o n s depend on t h e modulus and the u n i t w e i g h t o f t h e s o i l beneath t h e e x c a v a t i o n and behind the w a l l , the width and depth o f t h e e x c a v a t i o n * and the depth t o a firm layer. For s o i l s w i t h a high modulus, such as dense sands o r s o f t r o c k , t h e d e f o r m a t i o n s i n t h e s o i l beneath t h e e x c a v a t i o n l e v e l are i n s i g n i f i c a n t and t h e r e f o r e t h e a r c h i n g e f f e c t w i l l be s m a l l . I n these cases, the sum o f t h e s t r u t loads w i l l be n e a r l y equal t o t h e t h e o r e t i c a l Rankine v a l u e . However, t h e d i s t r i b u t i o n o f e a r t h pressure w i l l approach t h e shape as shown i n Figure 28(a) : t h e e a r t h pressure on t h e upper p a r t o f the w a l l i s only s l i g h t l y i n excess o f t h e t h e o r e t i c a l Rankine v a l u e and the arching e f f e c t . i s mainly due t o l o c a l r e d i s t r i b u t i o n o f e a r t h pressure between t h e lowest s t r u t and t h e excavation l e v e l ( B j e r r u m e t a l , 1972). T h i s i s a l s o the case f o r excavations i n s o i l s w i t h a somewhat lower modulus i f the width o f t h e excavation i s small compared t o i t s d e p t h . For wide excavation i n s o i l s w i t h low modulus, such as s o f t c l a y s , medium s t i f f c l a y s , s i l t s and loose sands, t h e deformations beneath t h e e x c a v a t i o n l e v e l are l a r g e , the arching e f f e c t w i l l be s i g n i f i c a n t and loads w i l l be thrown onto t h e s t r u t s such t h a t t h e sum o f t h e s t r u t loads w i l l c o n s i d e r a b l y exceed the t h e o r e t i c a l Rankine v a l u e . The e a r t h pressure d i s t r i b u t i o n w i l l be close t o t h e shape shown i n Figure 2 8 ( b ) . The design considerations f o r m u l t i - l e v e l s t r u t t e d w a l l s are t h e same as those f o r t h e case o f s i n g l e - l e v e l s t r u t t e d w a l l s .
7.2.2
Ultimate Limit States
The w a l l p e n e t r a t i o n below f i n a l excavation l e v e l should a l s o be s u f f i c i e n t t o l i m i t the bending moment i n t h e w a l l . A check s h o u l d be c a r r i e d out by considering t h e e q u i l i b r i u m o f t h e free-ended span below t h e owest s t r u t , assuming f i x i t y at t h a t s t r u t ( F i g u r e 2 9 ) . For t h e temporary ( c o n s t r u c t i o n ) c o n d i t i o n , a f a c t o r o f s a f e t y not l e s s than 1.5 on p a s s i v e resistance is usually a p p l i e d .
49
The depth of penetration of the wall below the excavation level should be sufficient to prevent foundation heave in cohesive soils (see Section 6.2.3) or hydraulic failure in cohesionless soils (see Section 6.2.4).
7.2.3
Serviceability Limit States
Ground movements around an excavation are usually estimated using the methods given in Section 9.3. However, the lateral deflection of a multilevel strutted wall is of secondary importance and deformation analysis is usually not carried out. If required, the 'beam on elastic foundation1 approach (see Sections 5 . 4 ) , the boundary element and finite element methods (see Sections 5.5 and 5.6) may by be used to predict the lateral displacement of the wall. A parametric study was carried out by Palmer & Kenney (1972) to investigate the behaviour of a strutted excavation with fully penetrating sheet pile walls through weak clay to bedrock. The results of their study showed that, of the parameters concerning soil conditions, the soil deformation modulus has the greatest influence. The influence of shear strength and soil/wall adhesion is of minor importance. This may be the case for an excavation with a fully-penetrating wall. However, in the case of a partially-penetrating wall, the other parameters might have greater influence. Potts & Fourie (1984) showed that the initial insitu stress has a considerable influence on the deformation of single-level strutted walls. It is considered that the same is true for multi-level strutted walls. All of the parameters concerning support conditions are important, but of these, wall stiffness is the most important. The effective strut stiffness depends not only on the stiffness of the struts and walings but also on the construction methods used to eliminate 'slack' between the wall and the walings and struts (such as wedging or other means of filling any gaps between the wall and the walings, and prestressing the struts). Where it is important to keep soil deformation, and especially surface settlement, to a minimum to prevent damage to neighbouring structures and public utilities, stiff sheet piles and a stiff support system (i.e. one with stiff walings and closely-spaced struts) have to be used. However, although these measures will be useful, some surface settlement is inevitable, primarily because of the amount of deformation which occurs below the bottom of the excavation. Prestressing of struts for an excavation in weak clay will only reduce deformation by a certain amount. Significant deformation will still occur, and increased strut loads must be anticipated. It is suggested that a small amount of prestress could be beneficial in ensuring the effectiveness of the support system at an early stage. However, a large amount of prestress may not provide additional benefits (see Section 9.2.7). The vertical spacing of struts affects the magnitude of strut loads and also of wall deflections, particularly where a flexible wall is used. A close horizontal strut spacing could provide a significant contribution to the effective stiffness at a particular strut level. A stiff waling has a similar effect.
50
7.2,4
Earth Pressures f o r Structural Design
The s t r u c t u r a l design o f m u l t i - l e v e l s t r u t t e d w a l l s i s u s u a l l y based on t h e s e m i - e m p i r i c a l a p p a r e n t e a r t h pressure envelopes developed by T e r z a g h i & Peck ( 1 9 6 7 ) , w h i c h was l a t e r summarised by Peck (1969b) (Figure 30)* The f o l l o w i n g f o u r important f e a t u r e s o f t h e T e r z a g h i & Peck method should be noted (Lambe, 1970) : t o a 'deep 1 e x c a v a t i o n , deep b e i n g "depths g r e a t e r than about 6 m (20
(a)
I t applies d e f i n e d as feet)".
(b)
The water t a b l e i s assumed t o be below t h e bottom o f t h e e x c a v a t i o n . Sand i s assumed t o be d r a i n e d , w i t h zero pore p r e s s u r e ; clay i s assumed t o be undrained and o n l y t o t a l stresses a r e c o n s i d e r e d .
(c)
The apparent pressure diagrams are not i n t e n d e d t o r e p r e s e n t t h e a c t u a l e a r t h pressure or i t s d i s t r i b u t i o n w i t h d e p t h , but r a t h e r envelopes from which safe estimates o f t h e s t r u t loads can be derived.
(d)
The behaviour o f an excavation i n c l a y depends very much on a s t a b i l i t y number, N c , where N c = H ) / C u . " E x t r a o r d i n a r y l a r g e movements appear when t h e depth o f a c u t i n c l a y approaches t h a t c o r r e s p o n d i n g t o N c = 6 o r 7 and when a considerable depth o f s i m i l a r c l a y a l s o extends below t h e bottom o f tire e x c a v a t i o n " (Peck, 1969b).
The Canadian F o u n d a t i o n Engineering Manual (Canadian G e o t e c h n i c a l S o c i e t y , 1985} has recommended t h a t h y d r o s t a t i c a l l y d i s t r i b u t e d water pressures should be added t o t h e apparent pressure envelopes f o r sands and s t i f f t o hard f i s s u r e d c l a y s , and t h a t f o r s o f t t o f i r m c l a y s an apparent earth pressure smaller than t h e h y d r o s t a t i c water pressure should not be used. For e x c a v a t i o n s i n deep d e p o s i t s o f s o f t c l a y , Henkel (1971) proposed a r a t i o n a l method f o r c a l c u l a t i n g t h e r e s u l t a n t t h r u s t on t h e s i d e s o f t h e e x c a v a t i o n as an a l t e r n a t i v e t o t h e apparent p r e s s u r e envelopes (Figure 3 1 ) . There are also other s e m i - e m p i r i c a l methods o f o b t a i n i n g h o r i z o n t a l s o i l s t r e s s e s and s t r u t l o a d s , e . g . t h e methods g i v e n i n Teng ( 1 9 6 2 ) , Armento ( 1 9 7 2 ) , T s c h e b o t a r i o f f (1973) and t h e Japan S o c i e t y o f C i v i l Engineers ( 1 9 7 7 ) . The a p p a r e n t p r e s s u r e e n v e l o p e s g i v e n i n t h e s e references are generally c o n s e r v a t i v e , but they are less w i d e l y used than those o f Terzaghi & Peck. Swatek e t al ( 1 9 7 2 ) , however, found reasonable agreement w i t h t h e T s c h e b o t a r i o f f pressure envelopes i n d e s i g n i n g t h e s u p p o r t system f o r a 21 m deep excavation i n Chicago c l a y . I t appears t h a t t h e T s c h e b o t a r i o f f method may be more a p p r o p r i a t e when t h e e x c a v a t i o n depth exceeds about 16 m. I t s h o u l d be n o t e d t h a t none o f t h e s e e n v e l o p e s a r e based on measurements taken from s a p r o l i t e s , t h e m a t e r i a l p r o p e r t i e s o f which a r e s i g n i f i c a n t l y d i f f e r e n t from sedimented sands and c l a y s , and t h e mass o f which f r e q u e n t l y contains d i s c o n t i n u i t i e s t h a t act as p r e f e r e n t i a l f l o w
51
paths for g r o u n d w a t e r and often as potential sliding surfaces. Preferential flow paths have an important bearing on the design against piping and heave at the base of the excavations. On the other hand, the presence of corestones can reduce the magnitude of earth pressures acting on the walls of the excavation (Massey & Pang, 1989). The apparent pressure diagram derived by Massad (1985) and Massad et al (1985) for the Brazilian lateritic and weathered sedimentary soils show a maximum value of 0.08VH (Figure 32) where VH is the overburden pressure. This is considerably lower than the corresponding value in the Terzaghi & Peck envelope for stiff sedimentary soils of other origins. Wirth & Zeigler (1982) found that the measured strut loads for the Baltimore Subway strutted excavation in soils derived from insitu weathering of gneissgranite and schists were about 70% of the Terzaghi & Peck envelope for sands. Although there are no reported field measurements of strut loads for Hong Kong residual soils and saprolites, observations from the performance of existing structures suggest that active and at-rest pressures are smaller than would be estimated by conventional calculations (Lumb, 1977; Endicott, 1982; Howat, 1985). Thus, field measurements of strut loads should be carried out in order to establish the apparent earth pressure envelopes for residual soils, which could be much lower than those of Terzaghi & Peck. In the determination of strut loads from apparent pressure envelopes, a reaction at the base of the excavation is assumed to exist. This reaction is provided by the passive resistance of the soil beneath the excavation. Figure 33 illustrates the method given by Goldberg et al (1976) for determining the depth of penetration in 'competent1 soils (e.g. medium dense to dense granular soils and stiff to hard clays), which are capable of developing adequate passive resistance. In weak soils, such as soft clays, the passive resistance may never reach the value of the active pressure on the retained side, no matter how deep the penetration is. This may also apply to a deep excavation in competent soil where the pressure on the wall below the lowest strut level is in an active state due to a high lateral pressure resulting from surcharge loading. For such cases, Goldberg et al (1976) have recommended that the wall should be driven to a depth required to prevent bottom heave or piping, and be designed as a cantilever about the lowest strut level. This approach is similar to that given in NAVFAC (1982b) for limiting the bending moment in the wall (see Section 7.2.2). Surcharge loadings should also be included in the apparent pressure envelopes to determine the strut loads. Support systems for deep excavations in urban areas are often subjected to asymmetric surcharges. An asymmetric condition occurs when the foundations of the buildings on the two sides of an excavation are of different types or when the buildings have a different number of floors or are at different distances from the excavation. Another condition of asymmetry is when the water table has been lowered more on one side of the excavation than the other. Analysis and field measurements by De Rezende Lopes (1985) showed that the wall opposite to the higher surcharges could be subjected to higher internal forces and could even be forced against the retained ground, developing passive pressure. Hence, both sides of the support system should be considered when the excavation is subjected to asymmetric surcharges. All construction stages of a strutted excavation should be considered, including strut removal stages. The required penetration depth, and the
52
s e c t i o n s t o be used f o r t h e s t r u t s , w a l i n g s and t h e w a l l s h o u l d be adequate f o r a l l c o n s t r u c t i o n stages. A step by s t e p procedure f o r designing s t r u t t e d excavations i s i l l u s t r a t e d by an example g i v e n by NAVFAC (1982b), Scott et al (1972) showed t h a t t h e e a r t h p r e s s u r e d i s t r i b u t i o n o f a s t r u t t e d excavation i n dense sand can be approximated by t h e extended Dubrova's s o l u t i o n (Dubrova, 1963) f o r a given w a l l d e f o r m a t i o n . The four p o s s i b l e t y p e s o f d e f o r m a t i o n o f a s t r u t t e d e x c a v a t i o n f o r which a g r a p h i c a l s o l u t i o n have been g i v e n a l l r e s u l t i n p a r a b o l i c l a t e r a l earth pressure d i s t r i b u t i o n c u r v e s . An e n p i r i c a l method f o r t h e design o f m u l t i - t i e d diaphragm w a l l s was f i r s t i n t r o d u c e d by L i t t l e j o h n et a l (1971) and was p r e s e n t e d again by James & Jack (1975) (see Section 8 . 2 . 4 ) . I t i n v o l v e s a procedure which c a l c u l a t e s t h e p o s i t i o n and magnitude o f a r e s u l t a n t t i e a t any s t a g e of excavation by t r e a t i n g the w a l l as a s i n g l e - t i e d s t r u c t u r e . I t has the advantage o f being a r e p e t i t i v e s i n g l e - t i e d w a l l design and i s amenable t o varying s o i l layers. Wood (1983) developed a computer program u s i n g t h i s method f o r t h e design o f m u l t i - s t r u t t e d diaphragm w a l l s . Bowles (1988) used a d i s c r e t e element method t o s i m u l a t e t h e s t a g e by stage c o n s t r u c t i o n o f a s t r u t t e d e x c a v a t i o n . Coulomb/Rankine t h e o r y was used t o c a l c u l a t e t h e e a r t h pressure above t h e e x c a v a t i o n l e v e l , and t h e concept o f subgrade r e a c t i o n was used f o r o b t a i n i n g t h e p a s s i v e r e s i s t a n c e below t h e excavation l e v e l . M a f f e i et a l (1977b) proposed a model f o r analysing m u l t i - l e v e l s t r u t t e d w a l l s which u t i l i s e s t h e e a r t h p r e s s u r e d i s p l a c e m e n t r e l a t i o n s h i p o f t h e s o i l , as w e l l as i n v o k i n g W i n k l e r ' s hypothesis ( W i n k l e r , 1867). A l l o f these 'beam on e l a s t i c f o u n d a t i o n 1 approaches can at best g i v e approximate values f o r t h e s t r u t l o a d s , w a l l bending moments and w a l l displacements (see S e c t i o n 5 . 4 ) . Both boundary element methods and f i n i t e element methods were used by F i t z p a t r i c k et a l ( 1 9 8 5 ) , Humpheson et a l (1986) and Wood & P e r r i n ( 1 9 8 4 ) , and by C l a r k e & Wroth ( 1 9 8 4 ) , E i s e n s t e i n & Medeiros ( 1 9 8 3 ) , Fernandes (1985), G a r r e t t & Barnes ( 1 9 8 4 ) , Hubbard e t a l ( 1 9 8 4 ) , Izumi e t a l ( 1 9 7 6 ) , Palmer & Kenney ( 1 9 7 2 ) , Roth et a l (1979) and Tominaga. e t - a l (1985) r e s p e c t i v e l y t o p r e d i c t t h e performance o f s t r u t t e d e x c a v a t i o n s . Q u a l i t a t i v e agreement between p r e d i c t e d and measured e a r t h p r e s s u r e d i s t r i b u t i o n s was o b t a i n e d . Good q u a n t i t a t i v e agreement, however, r e q u i r e s the choice o f s u i t a b l e values o f deformation p a r a m e t e r s . Both E i s e n s t e i n & Medeiros (1983) and Tominaga e t al (1985) used Lambe's (1967) s t r e s s path method t o determine t h e d e f o r m a t i o n p a r a m e t e r s . E i s e n s t e i n & Medeiros found t h a t t h e a n a l y s i s , which used a s o i l s t r e s s - s t r a i n model based on plane s t r a i n t e s t r e s u l t s , d e s c r i b e s t h e f i e l d behaviour more c l o s e l y than those based on conventional t r i a x i a l and s t r e s s - p a t h t r i a x i a l t e s t d a t a . The c a l c u l a t e d l a t e r a l pressure d i s t r i b u t i o n was a l s o found to d i f f e r from t h e s e m i - e m p i r i c a l pressure d i s t r i b u t i o n diagrams. E i s e n s t e i n & Negro ( 1 9 8 5 ) proposed a t h e o r e t i c a l framework f o r s t r u t t e d excavation a n a l y s i s ( F i g u r e 3 4 ) . Pressure i s uniquely r e l a t e d t o displacement Ah through two main c o n t r o l l i n g f a c t o r s : t h e a t - r e s t p r e s s u r e P o and t h e slope o f t h e ground r e a c t i o n curve. The former i s r e l a t e d t o t h e i n s i t u v e r t i c a l s t r e s s by K o , the c o e f f i c i e n t o f e a r t h pressure a t r e s t (see Section 5 . 6 ) , and t h e l a t t e r is a f u n c t i o n o f t h e s o i l ' s s t i f f n e s s and strength. The amount o f d i s p l a c e m e n t a l l o w e d i s a f u n c t i o n o f t h e excavation method and s u p p o r t i n g system used. In f a c t , the supporting system could be represented by a ' s u p p o r t r e a c t i o n curve 1 as shown i n
53 Figure 3 4 . At p o i n t 'A 1 i n t h i s f i g u r e , support and ground would i n t e r a c t and f i n d an e q u i l i b r i u m p o i n t , E. The c o n s t r u c t i o n technique would a f f e c t t h e amount o f d i s p l a c e m e n t , A h j , t a k i n g place p r i o r t o support installation. The w a l l and support s t i f f n e s s would be represented by t h e s l o p e , n, o f t h e normal t o t h e support r e a c t i o n curve shown. Thus, a f u n c t i o n c o u l d be p o s t u l a t e d t o represent t h e ground response : e
(~R> pr» §» o
Similarly,
K
o>
p^» s t r e n g t h , c o n s t r u c t i o n technique) = 0 o
.
.
(2)
t h e support response f u n c t i o n could be :
S ( ^ , p~, f , o
Jp
flip,
support s t i f f n e s s )
=0
(3)
These functions could be obtained by numerical modelling, using an appropriate constitutive relationship for the soil. Eisenstein & Negro then applied this framework to explain that the reason for the low values of the lateral pressures observed in the Brazilian lateritic and sedimentary soils (Massad, 1985; Massad et al, 1985) are due to their lower K o values. Since the construction technique has a significant influence on the behaviour of strutted excavations, any application of the finite element method to practical cases should be used in parallel with Peck's (1969a) observational procedure, and allowance should be made for re-analysis to be performed so as to accommodate soil parameter re-estimates and construction changes. The semi-empirical methods do not consider soil-structure interaction and are conservative, as they generally overestimate the strut loads. The analytical methods which take into account soil-structure interaction have a better prospect of giving strut loads closer to reality. Maffei et al (1977a) compared the 'beam on elastic foundation1 approach (Maffei et al, 1977b) with the finite element method and found that the lateral stresses on the wall are not very sensitive to the coefficient of subgrade reaction, but the wall displacements are. In terms of professional time and cost to perform an analysis, the finite element method is the most expensive, and the 'beam on elastic foundation1 approach is cheaper than the boundary element method. If the ground displacement is the crucial factor, the finite element method is the only method capable of making a reasonable prediction.
7.2.5
Design Mater Pressures
Considerations relating are given in Section 6.5.
7.3
to the evaluation of design water pressures
TEMPERATURE EFFECTS
T e m p e r a t u r e changes may cause s i g n i f i c a n t changes i n s t r u t l o a d s . Chapman et a l (1972) showed t h a t , by using e l a s t i c t h e o r y , t h e d e f o r m a t i o n modulus o f t h e s o i l behind a s t r u t t e d w a l l can be estimated by comparing s t r u t - l o a d changes w i t h w a l l d e f l e c t i o n s f o r a given t e m p e r a t u r e change.
54
By combining t h e r e l a t i o n s f o r e l a s t i c displacement o f a s t r u t t e d w a l l with t h e e f f e c t o f t e m p e r a t u r e on l o a d and d i s p l a c e m e n t o f a s t r u t , the f o l l o w i n g approximate r e l a t i o n was obtained : AP = A s E s aAT/(l+3nA s E s H/A w EL)
.
(4)
where AP = load change due t o a temperature change o f AT, A s = c r o s s - s e c t i o n a l area o f
strut,
E s = Young's modulus o f s t e e l , a
= thermal c o e f f i c i e n t o f expansion f o r s t e e l ,
nAs = t o t a l area o f s t r u t s a c t i n g a g a i n s t t h e w a l l o f area A w , H - height o f e x c a v a t i o n , L = length of
strut,
E = Young's modulus o f
soil.
Chapman e t a l (1972) showed t h a t t h e above r e l a t i o n p r o v i d e s a reasonable means f o r e s t i m a t i n g t h e s t r u t load changes which r e s u l t from temperature changes. S t r u t load changes w i l l approach t h e load changes f o r a r e s t r a i n e d s t r u t i f t h e s o i l modulus i s h i g h , t h e t o t a l area o f t h e s t e e l s t r u t s d i v i d e d by t h e area o f t h e s t r u t t e d w a l l i s l o w , and t h e l e n g t h o f t h e s t r u t i s l a r g e i n comparison w i t h t h e h e i g h t o f t h e e x c a v a t i o n . For a p e r f e c t l y r e s t r a i n e d s t r u t , t h e i n c r e a s e i n s t r u t temperature change i s g i v e n by : AP = A s E s a A T
load due t o a
(5)
Load changes r e s u l t i n g from l a r g e t e m p e r a t u r e i n c r e a s e s s h o u l d be added to t h e s t r u t loads i f t h e apparent e a r t h p r e s s u r e s are based on d a t a o b t a i n e d f o r r e l a t i v e l y c o n s t a n t t e m p e r a t u r e s , o r s m a l l changes i n temperature. The t e m p e r a t u r e e f f e c t on s t r u t l o a d s has a l s o been discussed by Goldberg et a l (1976) and Massad ( 1 9 8 5 ) .
55
8.
8.1 8.1.1
TIED-BACK OR ANCHORED WALLS
SINGLE-LEVEL TIED-BACK WALLS General
The f a c t o r s which i n f l u e n c e t h e behaviour o f t i e d - b a c k or anchored f l e x i b l e w a l l s were reviewed by Bjerrum et a l (1972) and are discussed i n Section 7 . 1 . 1 . For anchored w a l l s w i t h s l o p i n g anchors, the e f f e c t o f t h e v e r t i c a l component o f t h e anchor f o r c e should be considered i n design (Figure 35). This v e r t i c a l f o r c e may cause p o s s i b l e downward movement o f t h e w a l l when t h e w a l l has not been d r i v e n t o r e f u s a l and t h e p o i n t r e s i s t a n c e i s low (see S e c t i o n 8 . 1 . 2 ) . I n t h e design o f anchored geotechnical consideration : (a) (b) (c) (d)
walls,
there
are
four
primary
areas
of
ultimate limit states, serviceability limit states, earth pressures for structural design, and design water pressures.
The consideration of ultimate limit states rather than serviceability limit states usually governs the design.
8.1.2
Ultimate Limit States
(1) Overall Stability. Possible failure mechanisms of anchored sheet pile walls are shown in Figure 36. The penetration depth is not sufficient in Figure 36(a) to resist the axial force (due to the anchor) acting on the wall. The vertical component of the load in the inclined anchor forces the sheet piles downwards into the underlying soil. The wall will also move outwards because of the inclination of the anchors. The sheet pile wall in Figure 36(b) rotates about the anchor level when the toe of the wall moves outwards. The total stability of the cut is not sufficient in Figure 36(c). Failure occurs along a slip surface which passes below the wall. The failure mechanism in Figure 36(d) is caused by rupture of the anchors or of the walings. The sheet pile wall will, in this case, rotate or fold around a point located just below the bottom of the excavation. This mechanism may develop even when a single anchor rod or strand ruptures. This is because arching is unlikely to assist in re-distributing the lateral earth pressure, unless the anchors are exceptionally close; a waling, if present, may assist. In any case, the load increase in adjacent anchors must equal the load lost in the failed anchor. Rupture of anchors is most likely to occur when anchor loads are increasing generally, in which case additional capacity may not be available and progressive failure may then take place. Although not specifically mentioned by Broms & Stille (1976), bond failure (anchor pull-out) is another mechanism which can be considered to have the same effects.
56
The flexural resistance of the sheet piles has been exceeded in Figure 36(e). One or several plastic hinges develop at the levels where the positive or negative bending moments reach a maximum. At failure, the wall folds around these hinges, Broms & Stille (1976) reported failures of anchored sheet pile walls in the soft clays of Sweden. The failure mechanisms shown in Figures 36(a), 36(b) and 36(c) predominate. These indicate the importance of considering the vertical support of anchored sheet pile walls and other possible factors, such as reduction of shear strength of the clay due to excess pore water pressures caused by pile driving nearby, which can decrease the axial load carrying capacity of the sheet pile. The failure mechanism shown in Figure 36(d) has also occurred. Failure of the anchors was caused principally by the lateral expansion of the soil due to frost action, and was seldom caused by stress corrosion of the anchor bars or by slippage of the anchor bolts or the wedges. The latter generally occurred during the development stage of new anchor systems. The failure mechanism shown in Figure 36(e), caused by yielding of the walings or of the sheet piles, has not occurred in Sweden and also appears to be rare in other parts of the world. This indicates that the methods currently used for the design of walings and sheet pile walls are safe. Daniel & Olson (1982) reported an anchored bulkhead failure in stiff fissured clays due to a lack of understanding of soil behaviour under conditions involving significant excavation near the bulkhead. The design was- adequate for undrained conditions, but was inadequate for fully drained conditions. In Hong Kong, there is no need to consider frost effects. The design of anchors and the structural design of sheet pile walls are discussed in Section 8.1.4. The deep seated failure in Figure 36(c) can be analysed using appropriate slope stability analyses, such as those discussed in the Geotechnical Manual for Slopes (GCO, 1984). The toe equilibrium preferred to applying the
failure mechanism, Figure 36(b), can be assessed using limit methods of analysis, and the free-earth support method is the fixed-earth support method. The different methods for factor of safety are discussed in detail in Section 6.2.2.
Rowe (1952) derived a flexibility number, p , for sheet pile walls :
n
where L
~
L
;•...-..••.
(6)
= length of sheet pile,
E w = modulus of elasticity of sheet pile material, I w = moment of inertia of sheet pile cross-section. It should be noted that p is not dimensionless, and is not applicable to cohesive soil. The values of p. which delineate 'stiff1 piles (whose bending moments can be calculated reasonably accurately using the freeearth support method) from "flexible1 piles (whose bending moments are smaller than those calculated using the free-earth support method) are given by log P = -3.5 and log P = -4.125 for loose sand and dense sand
57
respectively. These values, however, are in the imperial units adopted by Rowe (viz. feet for L, lbf/in2 for E w and in4/ft for I w ) . Browzin (1981) developed a dimensionless flexibility number, p^t for cohesive frictional soils : 2c
'd ~ T ^ - d
"K
vKK
a
where V = effective unit weight of soil, B - width of sheet pile, c = cohesion of soil in terms of total stress, K a = coefficient of active earth pressure.
loose sheet sheet sheet
For the free-earth support conditions, p^ must be less than five soils and less than one for dense soils. The limiting length for piles can be calculated from these limiting values of p^. When pile length is greater than the limiting value, the bottom end of pile can be considered to be fixed.
for the the the
The stability against penetration failure (Figure 36(a}) is adequate if the side resistance and reaction at the base or toe of the wall is sufficient to overcome the vertical component of the anchor forces. Browzin (1977) showed that, for walls with inclined anchors in cohesion less soils, the required driving depth is nearly the same as for walls with horizontal anchors. Also, the skin friction capacity of the piles is much higher than the shear stresses generated by the vertical component of the anchor load. Browzin (1982) subsequently showed that the condition of sinking of piles in cohesive soils is not critical because of the lower anchor forces resulting from the great depth of penetration that is needed for stability against overturning. Anchored walls with sloping anchors in cohesive soils should be designed to avoid the following three conditions that involve a single stability number, 4c/FH : (a) (b) (c)
overturning, sinking of the piles with sloping anchors, and heave at the bottom of the excavation (see Section 8.1.2(2)}.
A graph presented by Browzin (1981) shows the relationship between the stability number and the parameter M 1 + H/d (where H is the retained height of the excavation and d is the depth of penetration) for the above three conditions of equilibrium (Figure 37). From the. graph, which was derived for an essentially cohesive soil (0=O°~1O°), it can be seen that stability against overturning requires larger values of 4c/TH than stability against sinking except when # = 0°, where the lines representing overturning and sinking conditions intersect at (3 = 3.7. This leads to the conclusion that stability against overturning is the governing condition for anchored walls with sloping anchors. The equation developed for stability against overturning also shows the high sensitivity of the required depth, d, to the assumed angle of internal friction, 0, a fact
58
also pointed out by Kovacs et al (1974). Browzin (1983) later presented universal design charts (Figures 38 to 40) for calculating the required depth of penetration for stability against overturning, using the free-earth support method. The depth of penetration is particularly sensitive to the assumed values of 0 and c. Nataraj & Hoadley (1984) developed a pressure diagram to reduce the computational effort in the design of anchored walls in sands using the free-earth support method.
(2) Foundation Heave and Hydraulic Failure, Considerations for foundation heave and hydraulic failure are discussed in Sections 6.2.3 and 6.2.4 respectively.
8.1.3
S e r v i c e a b i l i t y L i m i t States
D e f o r m a t i o n c o n s i d e r a t i o n s f o r anchored w a l l s are u s u a l l y of secondary importance and are s i m i l a r t o those f o r s i n g l e - l e v e l s t r u t t e d w a l l s , which are discussed in Section 7 . 1 . 3 .
8.1.4
Earth Pressures f o r S t r u c t u r a l Design
The c a l c u l a t i o n o f bending moments f o r anchored w a l l s is t h a t f o r s i n g l e - l e v e l s t r u t t e d w a l l s , which i s d i s c u s s e d 7.1.4(1).
similar to in Section
As discussed i n Section 8 . 1 . 2 ( 1 ) , bending f a i l u r e o f t h e w a l i n g s o r of sheet p i l e s r a r e l y occurs i n p r a c t i c e . This indicates t h a t the c u r r e n t l y used design methods are adequate. Similar to s t r u t t e d w a l l s , a f a c t o r of safety is applied to the design anchor f o r c e . Further allowances ( e . g . s a c r i f i c i a l t h i c k n e s s ) are r e q u i r e d i f c o r r o s i o n problems are expected. Anchors are u s u a l l y i n c l i n e d , and t h i s has t h r e e p o s s i b l e e f f e c t s : (a)
The value o f Ka i s increased due t o r e v e r s a l o f wall f r i c t i o n under working c o n d i t i o n s (however, a t f a i l u r e , t h e w a l l f r i c t i o n r e v e r t s t o t h e usual d i r e c t i o n , so no change needs t o be made t o the c a l c u l a t i o n o f Ka f o r o v e r a l l s t a b i l i t y ) .
(b)
Settlement o f increased.
(c)
The anchor t e n s i o n i s g r e a t e r than t h e h o r i z o n t a l prop f o r c e .
the
wall
and t h e
retained
soil
is
equivalent
A comprehensive treatment o f t h e design o f anchors i s given by Hanna ( 1 9 8 0 ) , B r i t i s h S t a n d a r d s I n s t i t u t i o n (1989) and Geospec 1 • Model S p e c i f i c a t i o n f o r Prestressed Ground Anchors (GCO, 1989). Factors o f s a f e t y f o r anchor design i n Hong Kong are given i n Geospec 1 .
59
8.1.5
Design Water Pressures
Considerations relating are given in Section 6*5.
8.2 8.2.1
to the evaluation of design water pressures
MULTI-LEVEL TIED-BACK WALLS General
The behaviour of multi-level tied-back walls is shown qualitatively in Figure 41 (Hanna, 1968), The use of inclined anchors causes vertical loading of the wall as construction progresses, and inadequate bearing capacity at wall base level can contribute to wall failure. As the wall is loaded, displacements of the wall relative to the retained soil mobilize friction on the embedded wall and strains within the soil mass. Vertical wall movements resulting from wall compression and settlement of the wall base are responsible for lateral movements of the wall system. Observed behaviour of tied-back wall systems (Clough, 1972b) suggests that tied-back walls generally yield smaller deformations than strutted walls. Finite element analyses by Clough & Tsui (1974) have shown that much of the difference in behaviour is due to the fact that tied-back wall systems are usually prestressed, while strutted wall systems are usually not, and that over-excavation of the temporary excavation platform, a problem associated with strutted walls, can lead to substantial increase in movement. The process of development of earth loadings for strutted and tiedback walls is fundamentally different. For a strutted wall, active earth pressure is partially mobilised behind the wall, and the loads which develop are primarily a function of soil type and strength. Because of the high stiffness of the struts and the sequence of excavation, the top of the wall becomes essentially fixed after the first strut is installed, and the wall rotates about the top strut as excavation proceeds. This leads to higher loads near the top of the wall than would be indicated by Rankine/Coulomb theory. A tie-back system, on the other hand, has two possible modes of behaviour, both of which differ from that of a strutted wall. In the first mode, the wall is prestressed with a resultant greater than that of the active or partially active condition. Under these circumstances the system responds more akin to a foundation slab on soil, and the earth loading on the wall is primarily dictated by the loading distribution assumed in the design prestress diagram and relative wall and soil stiffnesses. In the second mode, the tied-back wall is prestressed to lower levels, i.e. total resultant equal to or less than that of active conditions, and the loadings on the wall are developed in a manner similar to a strutted wall. However, the upward redistribution of loads which occurs in a strutted wall does not occur in the case of a tied-back wall because the tie-back elements are very flexible, and the top of the wall will continue to rotate outwards as excavation progresses. Thus, the loads developed resemble those of a conventional triangular earth pressure distribution, instead of those usually developed in a strutted wall. The design considerations for multi-level tied-back walls are the same as in the case of single-level tied-back walls.
60
8.2.2
Ultimate Limit States
The failure mechanisms in Figure 36 are also possible for multi-level anchored walls. The design of the anchors and the structural design of the wall are discussed in Section 8.2.4. The stability against the deep-seated failure shown in Figure 36(c) can be assessed using appropriate slope stability methods (GCO, 1984). The vertical component of the anchor forces for multi-level anchored walls could be large. Stability against penetration failure is therefore an important consideration. The side resistance and the reaction at the wall base must be adequate to balance the vertical anchor forces. For this reason, it is common for multi-level anchored walls to be founded on rock. Anderson et al (1983) reviewed four methods of checking the overall stability of anchored retaining walls, viz. the German method (German Society for Soil Mechanics and Foundation Engineering, 1980), the Ostermayer (1977) method, the French method (Bureau Securitas, 1972) and the method by Littlejohn et al (1971). Jhey found that all four design methods resulted in stable systems, both at the end of construction and after backfill loading. The determination of anchor prestress loads by the method by James & Jack (1975) was found to be based on erroneous earth pressure assumptions (see Section 8.2.4). More consistent behaviour was found in tests designed using the method by Littlejohn et al (1971) for overall stability, particularly on surcharge loading, indicating that a logarithmic spiral method may be the most appropriate to use. However, Schnabel (1984) commented that the German and French methods are wrong, as they assume the tie-back pulls on the soil without pushing on the wall. He considered that only the Ostermayer method is correct because it considers the tie-backs as internal forces in the soil wedge between the wall and the anchors. The method by Broms (1968) also recognizes this. A comprehensive review of methods of analysing overall stability of anchored retaining walls is given in British Standards Institution (1989). The depth of penetration of the wall below excavation level should be sufficient to prevent foundation heave (see Section 6.2.3) or hydraulic failure (see Section 6.2.4).
8.2.3
SerYiceability Limit States
As in the case of multi-level strutted walls, deformation analysis of multi-level tied-back walls is of secondary importance and is usually not carried out. If required, the 'beam on elastic foundation1 approach (see Section 5.4), the boundary element and finite element methods (see Sections 5.5 and 5.6) can be used to predict the lateral displacement of the wall. Maffei et al (1977a) found that the wall displacement is very sensitive to the assumed coefficient of subgrade reaction. In their parametric study of the behaviour of tied-back walls, Clough & Tsui (1974) found that for a wall founded on rock, the wall movement and the soil settlement behind the wall can be substantially reduced by increasing the prestress load, the wall rigidity and the tie-back stiffness. The use of higher prestress loads can eliminate the wall movement at the wall top but is not able to prevent wall movement near the bottom of the excavation. As
61
a result, soil settlements occur in all stages of the excavation. The decrease in wall movement is not in direct proportion to the increase in wall rigidity and tie-back stiffness. Stroh (1975) showed that the wall movement is more than twice as large where clay is present below the wall compared with the case where the wall is founded on rock, and that the wall movement decreases almost linearly as the length of anchors increases.
8.2«4
Earth Pressures for Structural Design
The earth pressures that act on an anchored wall depend on the wall stiffness relative to the soil, the anchor spacing, the anchor yield and the prestress locked into the anchors at installation. Two conventional design methods which ignore the effects of wall rigidity, excavation depth and tie-back spacing on the earth pressure distribution acting on the wall are outlined below. An empirical method first introduced by Littlejohn et al (1971) was presented again by James & Jack (1975). This method is also recommended by the Canadian Foundation Engineering Manual (Canadian Geotechnical Society, 1985). Wood (1983) developed a computer program using this method (see Section 5.5). It can take account of the continuous wall construction, excavation and anchoring stages, and the procedure is applicable to varying soil layers. Certain basic assumptions are made in the method, e.g. the soil pressure distribution is assumed to be triangular and the wall is assumed to yield progressively as excavation proceeds. In addition, a point of contraflexure in the wall is assumed to occur at a location where the factor of safety against overturning is unity. Full-scale monitoring studies (Littlejohn & Macfarlane, 1975) have indicated that wall deflections and bending moments, which occur as excavation proceeds, follow a similar pattern to those predicted by this empirical design method. Moreover, the monitoring results suggest a triangular pressure distribution for a semi-rigid anchored diaphragm wall, which is different from the trapezoidal distribution assumed in the design of strutted excavations. Anderson et al (1983) found from their laboratory scale tests that the determination of anchor prestress loads by the method of James & Jack (1975) is based on erroneous earth pressure assumptions. They suggested that the method should be modified so that only half of the passive pressure acts in front of the wall, and that the mean of the active and atrest pressures acts behind the wall. The Canadian Foundation Engineering Manual (Canadian Geotechnical Society, 1985) suggests three possible values of earth pressure acting behind the wall : (a)
If moderate wall movements can be permitted, or where foundations of adjacent buildings extend to below the base of the wall, the pressure may be computed using the coefficient of active earth pressure K a .
(b)
If foundations of buildings or services exist behind the top of the wall at a distance smaller than the height of the wall and not closer than half the height, the pressure may be computed
62
using a coefficient K = 0.5(Ka + K o ) . (c)
If foundations of buildings or services shallow depth, at a distance smaller than height behind the top of the wall, the may be computed using the coefficient pressure at rest K o .
exist at half the pressure of earth
The Manual also suggests that the passive pressure should be reduced to account for the wall movement required to develop it. Another empirical method often adopted is the use of apparent pressure diagrams similar to the Terzaghi & Peck envelopes used in strutted excavations. A considerable variety of methods for calculating anchor prestress loads exists (Table 5 ) , and Figure 42 shows reported magnitudes of design prestress diagrams used in practice. Terzaghi & Peck envelopes for strutted excavations are also shown in the figure. It can be seen that, in most instances, prestressing loads applied to tied-back walls are higher than the upper limit values for strut loads. Laboratory-scale tests on rigid multi-anchored walls by Hanna & Matallana (1970), Hanna & Abu Taleb (1971, 1972), Plant (1972) and Hanna & Dina (1973) showed that a trapezoidal earth pressure distribution, with the earth pressure coefficient equal to 0.5(Ka + K o ) , is appropriate. For flexible multi-level tied-back walls, laboratory-scale tests by Hanna & Kurdi (1974) showed that a rectangular earth pressure distribution appears reasonable. dough & Tsui (1974) used a finite element method to investigate the behaviour of a tied-back sheet pile wall in clay. Their study showed that the calculated pressures are more triangular in shape than the design trapezoidal diagram, and the calculated pressures tend to concentrate slightly at the anchor levels, a phenomenon observed by Hanna & Kurdi (1974). The triangularization of these pressures is apparently caused by the movements of the flexible tie-back supports. Anderson' et al (1983) found from laboratory-scale tests that a trapezoidal earth pressure distribution assumption is realistic if excavation is being carried out to wall base level. However, a triangular pressure distribution is more realistic if excavation terminates some distance above wall base level. In practice, excavation would usually stop some distance above the wall base to minimize base movements, and field observations at the Guildhall Redevelopment in London (Littlejohn & MacFarlane, 1975) would seem to confirm the latter distribution. As discussed in Section 8.2.1, there are two possible modes of behaviour for tied-back walls, depending upon the prestress level employed. Ideally, prestress just sufficient to minimize settlements behind the wall should be used, bearing in mind that some settlement will always occur regardless of the prestress load, due to the unloading of the soil at the bottom of the excavation. Observed movements of actual tie-back systems suggest that the optimum level of prestress to minimize movements lies slightly above the load levels recommended by Terzaghi & Peck (1967) for strutted walls. Satisfactory performance can easily be obtained, however, using lower prestress levels if movements are not a primary concern J (dough, 1975). Hanna & Kurdi (1974) found that the maximum bending moment in a multi-
63
anchored w a l l decreases w i t h i n c r e a s i n g w a l l f l e x i b i l i t y . I f the w a l l height is taken t o be t h e f r e e span between two consecutive anchor l e v e l s , trends o f r e d u c t i o n i n bending moment w i t h f l e x i b i l i t y s i m i l a r t o Rowe's curves (Rowe, 1952) can be o b t a i n e d . Anderson et a l (1982) found from l a b o r a t o r y - s c a l e t e s t s t h a t i n o r d e r to r e s t r i c t movement o f anchored w a l l s s u p p o r t i n g b a c k f i l l w i t h a u n i f o r m s u r c h a r g e l o a d i n g , a r e c t a n g u l a r pressure d i s t r i b u t i o n , and an e a r t h pressure c o e f f i c i e n t o f Ko f o r t h e surcharge component appear r e a s o n a b l e . For a l i n e or s t r i p l o a d i n g , they suggested t h a t i n order t o r e s t r i c t movement, t h e s t r i p l o a d i n g component o f t h e r e c t a n g u l a r p r e s s u r e d i s t r i b u t i o n should be 1.3 times t h e maximum h o r i z o n t a l s t r e s s p r e d i c t e d by elastic theory. A n a l y t i c a l methods w h i c h t a k e s o i l - s t r u c t u r e i n t e r a c t i o n into c o n s i d e r a t i o n can be used t o determine t h e magnitude and d i s t r i b u t i o n o f e a r t h p r e s s u r e s a c t i n g on m u l t i - a n c h o r e d w a l l s . The 'beam on e l a s t i c f o u n d a t i o n 1 approach was a p p l i e d by Bowles (1988) and M a f f e i et a l ( 1 9 7 7 b ) , t h e boundary element methods by Wood (1979) and Pappin et a l ( 1 9 8 5 ) , and the f i n i t e element methods by Clough et al ( 1 9 7 2 ) , Clough & Tsui ( 1 9 7 4 ) , B a r l a & M a s c a r d i ( 1 9 7 5 ) , Murphy et a l ( 1 9 7 5 ) , Stroh & Breth ( 1 9 7 6 ) , Rosenberg e t a l ( 1 9 7 7 ) , Simpson et a l (1979), S t i l l e & Fredricksson (1979) and Hata et a l ( 1 9 8 5 ) , t o model the w a l l / a n c h o r / e x c a v a t i o n system, Maffei et al (1977a) found t h a t , using t h e 'beam on e l a s t i c f o u n d a t i o n 1 method, the h o r i z o n t a l stresses a c t i n g on t h e w a l l are not very s e n s i t i v e t o t h e c o e f f i c i e n t o f t h e subgrade r e a c t i o n when these s t r e s s e s are compared w i t h those o b t a i n e d by t h e f i n i t e element method. As i n . t h e case o f m u l t i - l e v e l s t r u t t e d w a l l s , t h e f i n i t e element a n a l y s i s (as w e l l as other a n a l y t i c a l methods) should be used i n p a r a l l e l w i t h an o b s e r v a t i o n a l p r o c e d u r e , and a l l o w a n c e ^ s h o u l d be made f o r r e - a n a l y s e s t o be p e r f o r m e d so as t o accommodate s o i l parameter r e - e s t i m a t e s and c o n s t r u c t i o n changes. Soil parameters a r e best r e f i n e d by comparing t h e observed performance a t e a r l y stages o f e x c a v a t i o n w i t h f i n i t e element p r e d i c t i o n s . Summarising, t h e a v a i l a b l e methods a r e , i n i n c r e a s i n g o r d e r o f c o s t o f p e r f o r m i n g an a n a l y s i s , t h e method o f apparent pressure diagrams, t h e s e m i e m p i r i c a l method by James & Jack ( 1 9 7 5 ) , t h e 'beam on e l a s t i c f o u n d a t i o n 1 a p p r o a c h , t h e boundary element and the f i n i t e element methods.
8.2.5
Design Water Pressures
Considerations relating to the evaluation of design water pressures are given in Section 6.5.
64
65
9.
9.1
GROUND MOVEMENTS AROUND EXCAVATIONS
GENERAL
This chapter considers only ground movements caused by deep excavations. An excellent review of movements of excavation support systems in soft clay is given by Clough & Schmidt (1981). Ground movements caused by pile driving and dewatering are discussed in the state-of-the-art report by D'Appolonia (1971)•
9.2 9.2.1
FACTORS INFLUENCING MOVEMENTS Effect of Stress Changes
Figure 43 illustrates, in a qualitative manner, the stress changes and strains experienced by two soil elements near an excavation. The stress paths are typical for a normally consolidated plastic clay. Both the reduction in total vertical and horizontal stress, and the change in equilibrium pore water pressure, are important. The table in the figure lists the stress changes and the strains brought about by the excavation and the change in seepage conditions. The critical factor that determines the inward movement of the sides of the cut below excavation level, and hence the magnitude of settlement, involves the proximity of the unloading stress path _for element B to the failure envelope. If the effective stress points B\ and B s s , are well within the effective stress failure envelope, Kf, i.e. if local yield is avoided, then the heave will be small and the lateral movement of the soil below excavation level will be small. On the other hand, if the effective stress points for element B approach the failure envelope and passive local yield occurs, then the movements will be large.
9.2.2
Dimensions of Excavation
The deeper the excavation, the greater is the decrease in total stress, and thus the larger are the movements of the surrounding soil.
9.2.3
Soil Properties
In clays, it has been shown (Clough et al, 1979; Mana & Clough, 1981) that the maximum lateral wall movement could be correlated with the factor of safety against basal heave, which in turn depends on the shear strength of the soil. The rate and magnitude of movement increase rapidly as the factor of safety approaches one. . If a clay is presumed to be isotropic when it is in fact anisotropic, the basal heave factor of safety may be much lower and lateral wall movements as well as ground surface settlements may be larger than expected (Clough & Hansen, 1981). Wall movements and ground settlements are smaller in stiff soils, such as granular soils and stiff clays, than in soft soils, such as soft and medium clays and compressible silts (Peck, 1969b).
66
9.2.4
Initial Horizontal Stress
For excavated w a l l s i n s o i l s w i t h a high i n i t i a l K o v a l u e , l a r g e s o i l and wall movements are experienced even at s h a l l o w depths o f e x c a v a t i o n . The w a l l b e h a v i o u r i s d o m i n a t e d by v e r t i c a l u n l o a d i n g caused by the excavation p r o c e s s , and l a r g e movements s t i l l occur even i f the w a l l is f u l l y r e s t r a i n e d from h o r i z o n t a l movement. For w a l l s i n s o i l s w i t h a low Ko v a l u e , t h e d i s p l a c e m e n t s are much s m a l l e r i n magnitude ( P o t t s and F o u r i e , 1984).
9.2.5
Groundwater Conditions
In p r a c t i c e , a p e r f e c t l y w a t e r t i g h t w a l l p e n e t r a t i n g an impermeable s o i l l a y e r at t h e bottom o f t h e e x c a v a t i o n does not e x i s t . Where water flows i n t o t h e e x c a v a t i o n , a decrease i n groundwater p r e s s u r e w i l l occur. This w i l l cause an increase i n e f f e c t i v e s t r e s s and s e t t l e m e n t o f t h e s o i l surrounding the e x c a v a t i o n . Where groundwater has n o t been brought under complete c o n t r o l , l a r g e , e r r a t i c and damaging s e t t l e m e n t due t o t h e f l o w or the m i g r a t i o n o f f i n e s i n t o t h e excavation are not uncommon*
9.2.6
Stiffness of Support System
The support system stiffness depends on the stiffness of the wall and its supports, the spacing between supports, and the length of wall embedded below the excavation bottom. A valuable study into the effect of wall stiffness and support spacing was carried out by Goldberg et al (1976), and the results are presented in Figure 44, in which the stability number FH/C U is plotted versus the stiffness parameter E w I w /h 4 , where E W I W is the stiffness of the wall, h is the distance between supports, FH is the overburden pressure, and C u is the undrained shear strength of the soil. Various boundary lines are drawn to establish zones of expected lateral wall movements. These data suggest that an increase in E w I w /h 4 has a • significant effect in reducing movements. The movement is also a function of the factor of safety against basal heave, being more significant at lower factors of safety than at higher ones. Increasing the stiffness of the strut or the tie-back decreases movements, but this effect shows diminishing returns at very high values of strut or tie-back stiffness. Movements are also reduced as the depth to an underlying firm layer decreases and when the wall toe is embedded into the underlying firm layer.
9.2.7
Effect of Preloading
Preloading removes slack from the support system, and thus eliminates this potential source of movement. Each application of preload reduces the shear stresses set up in the soil due to previous excavation activities. This means that the soil is partially unloaded, and its stress-strain response is stiffened until the next excavation step generates shear stresses that reload the soil beyond its maximum previous level of shear stress. This temporary stiffening of the soil also leads to reduced movements.
67
Use of preloads in the struts or tie-backs reduces movement, although there are diminishing returns at higher preloads. Very high preloads may, in fact, be counter-productive, since local outward movements at support levels can damage adjacent utilities. Clough & Tsui (1974) showed that the prestress loads in tie-backs calculated in accordance with a triangular at-rest pressure diagram are less effective in reducing movements than those obtained from the trapezoidal pressure diagram of the shape recommended by Terzaghi & Peck (1967) for calculating strut loads for strutted excavations. Clough (1975) also found that the optimum effect of prestress loads in reducing system movements can be achieved by using levels slightly greater than those recommended by Terzaghi & Peck (1967). Model tests by Hanna & Kurdi (1974) have confirmed this conclusion. Clough & Tsui (1974) also showed that movements of the tied-back wall, and the soil settlements behind the wall, can be substantially reduced by a judicious choice of prestress load and support system stiffness. O'Rourke (1981) considered that, in most cases, cross-lot struts prestressed to 50% of their design load will be sufficiently rigid to restrict further movement at the level of the supports, and sufficiently low in load to avoid being overstressed as additional excavation occurs.
9.2.8
Effect of Passive Soil Buttresses
The effectiveness of these in controlling movements in excavation in soft to medium clay was studied by Clough & Denby (1977) using the finite element method. They used the term "berm11 for a passive soil buttress. Figure 45 shows the theoretical relationships between settlements behind bermed sheet pile walls and the stability number, ^H/C U 5, where C U 5 is the undrained shear strength at the base of the excavation, for the condition of excavation after the berm has been removed and the rakers installed. At low stability numbers, increases in berm size produce minimal movement reduction. For intermediate stability numbers, increases in berm size show a diminishing effect. At high stability numbers, increasing from an intermediate to a large berm leads to a large reduction in settlements. However, this reduction is an exarrple of false economy, since at large stability numbers, large movements occur even with large berms because deep-seated movements take place beneath the berms. As a result, the effectiveness of the berms is diminished.
9.2.9
Associated Site Preparation Works Site preparation may include the following activities : (a)
relocation and underpinning of utilities,
(b)
dewatering of aquifers above and below the base of the excavation,
(c)
construction of the excavation support system, and
(d)
the installation of deep foundations.
68 In some cases, the movements associated with the site preparation works will exceed those which occur as a result of the excavation and support process. The relocation of u t i l i t i e s , for example, may have a locally severe impact on an adjacent property, especially when trenching is carried out close to pipelines and communication conduits, Dewatering may consolidate the soil over an area which substantially exceeds the area affected by excavation-induced movements. Also, i t often causes settlements well in excess of excavation settlements. However, in areas which have been subjected to earlier dewatering activities, settlements due to further dewatering are smaller than those in virgin ground due to the stiffer response of the preconsolidated s o i l . Wall construction may require predrilling for soldier piles, the use of vibratory hammers to install sheet piles, or the excavation of slurry panels for concrete diaphragm walls. Each of these can cause permanent movements, the magnitude and distribution of which w i l l vary according to the soil conditions, site history and details of the construction procedures, 9.2.10 Influence of Construction Factors A review of the influence of construction factors on excavation movements was made by Clough & Davidson (1977), and useful specific case history accounts were provided by Broms & Stille (1976), Hansbo et al (1973), Lambe et al (1970), O'Rourke et al (1976), White (1976) and Zeevart (1972). Additional movements of excavations, and even local failure, have been produced by late installation of supports, over-excavation, pile driving, caisson construction, loss of water through holes for tie-backs and joints or interlocks of slurry or sheet pile walls, remoulding and undercutting of clay berms, and surcharge loads from spoil heaps and construction equipment. Lambe (1970, 1972) demonstrated that variations in wedging techniques between the waiings and struts, and differences in excavation procedure, can result in doubling of the wall and soil movements* Clough & Tsui (1974) used a finite element analysis to show that over-excavation could lead to a 100% increase in movement. Because these factors cannot always be quantified, i t is d i f f i c u l t to make accurate prediction of movements. However, many of the undesirable effects of these factors have been identified, and they can therefore be anticipated and controlled through good specifications, well-planned constructTon procedures and close supervision and monitoring. The designer must consider how the excavation and subsequent construction should be £n™ *«°7kf i d . e n t i f * critical construction factors and, where possible, allow for them in performance estimates and specifications. 9,3 PREDICTIVE METHODS 9.3.1 Types of Methods Prediction of soil movements behind a supported excavation can be made
69
by the following methods : (a) (b) (c) (d)
9.3.2 **•
e m p i r i c a l methods, s e m i - e m p i r i c a l methods, t h e method o f v e l o c i t y f i e l d s , and t h e f i n i t e element method.
Empirical Methods
S e t t l e m e n t behind excavations can be estimated based on data obtained from p r e v i o u s excavations in s i m i l a r s o i l conditions; Peck (1969b) summarised data on s e t t l e m e n t s behind excavations t o produce the u s e f u l c h a r t shown i n Figure 4 6 . Settlements presented as a percentage o f the e x c a v a t i o n depth were p l o t t e d a g a i n s t d i s t a n c e from the excavation d i v i d e d by e x c a v a t i o n d e p t h . Three d i f f e r e n t zones, as d e f i n e d i n t h e f i g u r e , were identified. G e n e r a l l y , where workmanship i s average or above average and s o i l c o n d i t i o n s are not e s p e c i a l l y d i f f i c u l t , s e t t l e m e n t s should not exceed one p e r c e n t o f t h e e x c a v a t i o n d e p t h . In cases where seepage can occur and s o i l c o n s o l i d a t i o n r e s u l t s , t h e one percent f i g u r e can be exceeded (Lambe et a l , 1970). I t should a l s o be noted t h a t c o n s t r u c t i o n technique can have a s t r o n g i n f l u e n c e on movements o f s t r u t t e d systems (see S e c t i o n 9 . 2 . 1 0 ) . O'Rourke e t a l (1976) compiled s e t t l e m e n t data associated w i t h s o l d i e r p i l e and l a g g i n g e x c a v a t i o n i n t h e dense sand and interbedded s t i f f clay o f W a s h i n g t o n , U.S.A. ( F i g u r e 4 7 ) . Expressed as a percentage o f t h e e x c a v a t i o n d e p t h , t h e s u r f a c e s e t t l e m e n t s were equal t o or less than 0.3% near t h e edge o f t h e cut and 0.05% a t a d i s t a n c e equal t o 1.5 times t h e e x c a v a t i o n d e p t h . A comparative study o f s t r u t t e d excavations i n t h e s o f t c l a y o f Chicago using temporary berms and r a k i n g s t r u t s was a l s o c a r r i e d out by 0'Rourke e t a l ( 1 9 7 6 ) . The data are p l o t t e d i n dimensionless form i n F i g u r e 48. Three zones o f ground displacement can be d i s t i n g u i s h e d and r e l a t e d to the s a l i e n t c h a r a c t e r i s t i c s of construction. These zones approximate t o t h e t h r e e zones o f s e t t l e m e n t d e l i n e a t e d by Peck ( 1 9 6 9 b ) , w i t h t h e e x c e p t i o n t h a t t h e widths o f t h e s e t t l e m e n t zones a r e n o t a b l y shorter. This i n d i c a t e s t h a t t h e s e t t l e m e n t s associated w i t h t h e Chicago e x c a v a t i o n s a r e c o n f i n e d t o areas t h a t are c o m p a r a t i v e l y nearer t h e edges of the excavations. Based on t h e s e s t u d i e s , 0 ' R o u r k e (1981) noted t h a t t h e s t r u t t e d e x c a v a t i o n s were g e n e r a l l y c a r r i e d out i n t h r e e prominent stages ( F i g u r e 49) : (a)
Initial excavation before strutting. The e x c a v a t i o n was deepened b e f o r e s u p p o r t s were i n s t a l l e d . Thus, d e f o r m a t i o n o f t h e w a l l occurred p r i m a r i l y as a c a n t i l e v e r - t y p e movement. The h o r i z o n t a l s t r a i n s p r o d u c e d i n t h i s mode o f d e f o r m a t i o n form a t r i a n g u l a r p a t t e r n o f contours t h a t decrease i n magnitude w i t h depth and d i s t a n c e away from t h e w a l l .
(b)
Excavation t o subgrade a f t e r i n s t a l l a t i o n o f upper s t r u t s . As t h e supports were i n s t a l l e d , t h e upper p o r t i o n o f t h e w a l l was r e s t r a i n e d from f u r t h e r l a t e r a l movement. I n t h e deeper p o r t i o n o f t h e e x c a v a t i o n , i n w a r d b u l g i n g o f t h e w a l l caused
70
t e n s i l e s t r a i n s , t h e c o n t o u r s f o r w h i c h were i n c l i n e d at approximately 45° t o t h e v e r t i c a l . (c)
Removal o f s u p p o r t s . As t h e bottom s t r u t s were removed t o b u i l d t h e u n d e r g r o u n d structure, f u r t h e r inward b u l g i n g o f t h e w a l l o c c u r r e d . When t h e upper s t r u t s were r e m o v e d , t h e w a l l was s u p p o r t e d i n i t s l o w e r p o r t i o n by t h e subway structure, resulting in a c a n t i l e v e r - t y p e d e f o r m a t i o n a t t h e upper p o r t i o n o f t h e w a l l . Consequently, the cumulative s t r a i n s were a composite o f the h o r i z o n t a l d i s t o r t i o n a s s o c i a t e d w i t h inward b u l g i n g a t depth and t h e d i s t o r t i o n associated w i t h c a n t i l e v e r movement a t t h e upper levels of excavation.
T h u s , t h e s o i l s t r a i n s are c l o s e l y r e l a t e d t o t h e modes o f w a l l deformation. A q u a n t i t a t i v e r e l a t i o n s h i p between t h e d i s p l a c e m e n t p a t t e r n of t h e w a l l and movement o f t h e ground s u r f a c e a d j a c e n t t o t h e e x c a v a t i o n was developed by O'Rourke (1981) and i s shown i n F i g u r e 50. A c o e f f i c i e n t of d e f o r m a t i o n , CQ, was defined as t h e r a t i o o f t h e c a n t i l e v e r movement of t h e w a l l , S w , t o t h e sum o f t h i s c a n t i l e v e r component and t h e inward b u l g i n g component, S w ' , o f t h e d i s p l a c e m e n t . I t s h o u l d be noted t h a t the data summarised in Figure 50 p e r t a i n t o excavations where t h e w a l l s have been d r i v e n t o a s t i f f s o i l l a y e r , and thus do not have s i g n i f i c a n t movement a t t h e bottom. The f i e l d data i n Figure 50 show t h a t t h e l i m i t s f o r t h e r a t i o o f h o r i z o n t a l t o v e r t i c a l movement ( s e t t l e m e n t ) are a p p r o x i m a t e l y equal t o 0.6 and 1.6 f o r w a l l deformation caused s o l e l y by inward b u l g i n g ( i . e . w i t h the w a l l f i r m l y s u p p o r t e d a t an e a r l y stage o f e x c a v a t i o n ) and s o l e l y by c a n t i l e v e r movement, r e s p e c t i v e l y . These bounds compare f a v o u r a b l y w i t h t h e l i m i t s determined from model t e s t s i n sands performed by M i l l i g a n (1983) (see Section 9 . 3 . 4 ) . Clough et a l (1979) and Mana & Clough (1981) reviewed case h i s t o r i e s o f s h e e t p i l e and s o l d i e r p i l e w a l l s i n clays s u p p o r t e d p r i m a r i l y by cross-lot struts. The r e s u l t s ' are summarised i n F i g u r e 5 1 , which shows t h a t t h e maximum l a t e r a l movement can be c o r r e l a t e d w i t h t h e f a c t o r o f s a f e t y a g a i n s t basal heave d e f i n e d by Terzaghi ( 1 9 4 3 ) . The movements increase r a p i d l y below a f a c t o r o f s a f e t y o f 1.4 t o 1 . 5 , w h i l e a t higher f a c t o r s o f s a f e t y the nondimensional movements l i e w i t h i n a narrow range o f 0.2% t o 0.8%. Moreover, t h e r e do not appear t o be any s i g n i f i c a n t d i f f e r e n c e s i n l a t e r a l movements between sheet p i l e w a l l s whose t i p s are embedded i n an underlying s t i f f l a y e r and those whose t i p s remain i n t h e moving clay mass. Mana & .Clough (1981) p l o t t e d maximum s e t t l e m e n t data versus l a t e r a l wairmovements as shown i n Figure 52. This f i g u r e shows t h a t s e t t l e m e n t s range from 0.5 t o 1.0 times the l a t e r a l w a l l movements. I f s e t t l e m e n t s a r e t 0 l a t ^"SS^JffSi^J1*11^ V b 6 1 ^ a l w a l l movements, then F i g u r e
197 ] summdr sed the t ^ / h ^ J } i 7 f n l J data from a v a i l a b l e l i t e r a t u r e on t i e d - b a c k w a l l movements, as shown i n Figure 53. Generally the w a l l movements and settlements are w e l l below one p e r c e n t T thef E x c a v a t i o n
71
depth, although in a few instances values above one percent appear. The data were re-plotted in Figures 54 and 55, where the percentage movement is plotted against the maximum ordinate of the design prestress diagram for sands and stiff clays respectively. The maximum prestress pressure is made nondimensional by the product of unit weight and excavation height. It can be seen that, for both sands and stiff clays, the movements decrease with increasing prestress. The design pressure levels suggested by Terzaghi & Peck (1967) are also shown in the figures. The data suggest that the optimum effect of prestressing in reducing movements is achieved by using pressure levels slightly greater than those of Terzaghi & Peck (1967)• Further data on movements of the crest and settlement of the ground behind tied-back walls are given in British Standards Institution (1989). A large amount of data have been obtained during the construction of the Mass Transit Railway (MTR) in Hong Kong. Stroud & Sweeney (1977) reported ground movements caused by the installation of a diaphragm wall test panel. Morton et al (1980a & b) found that the settlements induced by the MTR construction can generally be separated into those resulting from wall installation, dewatering and station box excavation. Settlements caused by the diaphragm wall construction were found to be significant. The reason for this was thought to be due to the breakdown of arching as the length of wall was constructed, and the high lateral swelling potential of the completely decomposed granite (Davies & Henkel, 1980). Cowland & Thorley (1984) also reported significant settlements due to slurry trench excavation up to a distance equal to the depth of the trench away. Significant settlements were measured at some buildings even when the whole of the base of their foundations was outside the theoretical active wedge. A relationship was found between maximum building settlement and the ratio of the foundation depth to trench depth (Figure 56). Dewatering was the prominent cause of building settlement at some of the MTR stations. Relatively large wall deflections occurred during excavation. However, the building settlements were usually low, the ratio of maximum lateral wall movement to building settlement being of the order *of 4:1. While the ground conditions and the causes of settlement of adjacent buildings differed from site to site, the total settlements recorded were, surprisingly, found to be related to their foundation depths as shown in figure 57. Settlement data associated with the construction of the MTR Island Line were compiled by Budge-Reid et al (1984). The settlements caused by diaphragm wall installation, station box dewatering and diaphragm box excavation are shown in Figures 58, 59 and 60 respectively. It should be noted that while Figures 56 and 58 give general empirical relationships for estimating settlement due to slurry trench installation, the fundamental factor affecting settlement is the effective stress change in the completely decomposed granite. Unfortunately, the only data relating settlement and the effective slurry pressure are those given by Davies & Henkel (1980). Very few settlement monitoring data associated with deep excavations for basement construction in Hong Kong have been published. Morton & Tsui (1982) presented very limited ground movement data for the construction of the deep basement of the China Resources Building. Well documented settlement monitoring data for the construction of the new Hong Kong Bank headquarters building were given by Fitzpatrick & Will ford (1985) and Humpheson et al (1986).
72
9.3.3 Semi-empirical Methods
The following semi-empirical methods have been proposed to relate the settlement of the ground behind an excavation with the lateral movement of the supporting wall. Caspe (1966) presented a method of analysis which relates the settlement profile behind an excavation with the wall deflection. Three key assumptions were made in the analysis : (a) There is a surface behind the wall which defines the limit of significant soil deformation, (b) The variation in horizontal strain in the soil at any level between the above surface and the wall is assumed. (c) At all points, the vertical strain is assumed to be related to the horizontal strain by the Poisson's ratio, p. Kane (1966) commented that the third assumption is incorrect, as the vertical strain should be related to horizontal strain by v/{l»v) under plane strain conditions. Consequently, the predicted settlement profile no longer matches the measured profile. Bowles (1988) modified Caspe1s method and obtained reasonable matching with the measured settlement profile. Another method was proposed by Bauer (1984) to obtain a reasonable estimate of ground loss around excavations in sands, and this is illustrated in Figure 61. 9.3.4 Method of Velocity Fields The method of velocity fields has been discussed in Section 5.3. Bransby & Milligan (1975) showed that the deformations in the soil behind a flexible cantilever retaining wall could be predicted from the correction of the wall using a simple velocity field. The method depends °J1 ? ^ l P?rameter> the angle of dilation, which can be related to the £ 9 l f f - i f r l c i l 2 I ! u sd1nng?l e tohfe dsi t1r ae tsi so n di latancy relationship of Rowe (1962). t0 wi hi fir S i !• i * n 50 is sufficiently accurate for ail practical purposes. l 1 * Wit h! he dnP d td nt ce h o r o of r d1su PP° rt nea r the top, Milligan (1983, numJL nf . a i ! + -1 V ei l - ? *Placements could be divided into a ?!L 1 J J /i simple zones, which could be related to the t h S e 20nes a r e l i n k e d in S c h arrMnn C0U",V* ^ t ^ * ™re complex zone without rha™ J d O c ? u r - Ford tSh iem pc1a es e vi en l 0w h i c h t h * ^oi 1 is deforming ^ ^ i t y f i e l d could apply, ive of the mode of deformation of the wall. 9.3.5
Finite Element Method
f t ^ s " ^ ^ ^ only the finite element takes account of the interaction between all the components within
73
the retaining system (Institution of Structural Engineers, 1989}• The inherent assumptions that have to be made in applying the finite element method to the prediction of ground movements were discussed by Borland (1978). The soil parameters required for a finite element analysis are discussed in Section 5.6. Bur land et al (1979) back-analysed measurements of movement using finite element techniques to provide insitu deformation parameters for London Clay, and used the derived parameters successfully in predicting and controlling ground movements around major excavations in critical locations in London. Recent field and laboratory tests reveal that some soils exhibit a higher stiffness at small strain (see Section 5.6). Jardine et al (1986) used the finite element method incorporating a soil model that features small strain nonlinearity to predict the settlement behind a strutted excavation. Good agreement with the observed field behaviour was obtained. Potts & Fourie (1984) used a finite element analysis which incorporated an elasto-plastic soil model, to investigate the behaviour of a single propped retaining wall. The results of their study showed that, for excavated walls in soils with a high initial Ko value, large soil movements occur even at shallow depths of excavation. This behaviour is related to the vertical unloading due to the excavation process. Large movements still occur even if the wall is fully restrained from horizontal movement. For walls in soils with a low K o value, the soil movements are much smaller in magnitude. Mana & Clough (1981) used a finite element model to verify the trend of field data, which indicate that there is a well-defined relationship between the maximum lateral movement and the basal heave factor of safety (Figure 62). The range of field data in Figure 51 are superimposed onto Figure 62, and these nicely bound the finite element trend curve. The results of the finite element analysis also showed that the maximum ground settlements range from 0.4 to 0.8 times the maximum lateral wall movements (Figure 6 3 ) . However, the field data showed that the maximum ground settlements were from 0.5 to 1.0 times the maximum lateral wall movements (Figure 52). The slightly larger ratio of settlement to lateral movement observed for some of the field cases is probably due to the effects of small amounts of consolidation, or traffic or surcharge loading behind the wall. Based on their finite element studies, Mana & Clough (1981) proposed the following procedures for estimating wall movements and ground surface settlements for a strutted excavation in soft to medium clays : (a)
Compute the factor of safety against basal heave using Terzaghi's approach at each construction stage where movements are desired. The factor of safety at each stage is to be used in the prediction procedures, except in the case where the factor of safety increases at later excavation stages due to the presence of an underlying strong layer. In this instance, the key factor of safety is the minimum which develops during excavation.
74
fb)
Estimate the maximum l a t e r a l w a l l movements A h m a x between the f a c t o r o f s a f e t y and w a l l movements, using the r e l a t i o n s h i p i n Figure 62. The maximum ground s e t t l e m e n t Avma
(c)
Based on the w a l l s t i f f n e s s f a c t o r , E w I w / P h 4 , t h e s t r u t s t i f f n e s s f a c t o r , S k / F H , t h e depth t o f i r m layer E^, and the excavation w i d t h B, d e t e r m i n e the i n f l u e n c e c o e f f i c i e n t s a\\, or$» a p » and a e » using Figures 64, 65, 66 and 67 r e s p e c t i v e l y .
(d)
Determine the i n f l u e n c e c o e f f i c i e n t design s t r u t preloads using Figure 6 8 .
(e)
S e l e c t a v a l u e f o r t h e modulus m u l t i p l i e r M ( = E / C U ) , based on a v a i l a b l e t e s t d a t a o r on i n f o r m a t i o n found i n t h e l i t e r a t u r e Then d e t e r m i n e the modulus m u l t i p l i e r influence c o e f f i c i e n t , aw> from Figure 69.
(f)
Using the value o f Ah^ d x obtained i n s t e p (b) and t h e i n f l u e n c e c o e f f i c i e n t s determined i n steps ( c ) , ( d ) , and ( e ) , a r e v i s e d value f o r t h e maximum l a t e r a l movement i s estimated as :
Ahjfiax
= Ah
max <*W a $ c*D a B a ? a M
.
.
for
the
• . (8)
(g) A revised estimate for the maximum ground settlement is. obtained by assuming that it is equal to 0.6 to 1.0 times Ahjfjax. (h) The ground settlement profile is established using the value of the maximum settlement from step (g) and Figure 70. •-dough & Tsui (1974) carried out finite element analyses to study the difference in behaviour between tied-back and strutted walls in clay. The1 results of their study showed that, in the case of walls with "equivalent conditions, the tied-back wall moved more than the strutted wall, and settlements behind the tied-back wall were also larger. This was attributed to the movement of the anchorages in the retained soil mass. However, in practice, the tied-back wall is usually prestressed and the strutted wall is not, and over-excavation often occurs during construction of the strutted wall. In these cases, the finite element analyses showed that the tied-back wall moved less and ground settlements were smaller than those for the strutted walls. Although the method of Mana & Clough (1981) for predicting movements is. primarily designed for cross-lot strutted walls, i t can be used in its present form for tied-back walls, provided that the anchors are embedded in an unmoving soil or rock mass. Should the anchorages be subjected to movement, the method is no longer applicable, since this will introduce additional displacements that are not considered in the method.
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10.
REVIB* OF LOCAL DESIGN PRACTICE
Local guidance on t h e design o f excavation support systems and on p r e d i c t i o n o f ground movements around excavations is given i n Chapters 7 and 10 o f Geoguide 1 : Guide t o R e t a i n i n g W a l l Design (GCQ, 1982) respectively. These chapters o f t h e Geoguide are based heavily on overseas p u b l i c a t i o n s from t h e 1 9 7 0 ' s . Some o f t h e major f a c t o r s which i n f l u e n c e d e s i g n , c o n s t r u c t i o n and c o n t r o l o f excavations f o r basement c o n s t r u c t i o n i n urban areas o f Hong Kong are reviewed by Sivaloganathan ( 1 9 8 6 ) . S t r u t t e d e x c a v a t i o n s a r e u s u a l l y d e s i g n e d , or at l e a s t they are checked, u s i n g t h e apparent pressure diagrams g i v e n by Peck (1969b) and t h e Japan S o c i e t y o f C i v i l Engineers (1977). These pressure diagrams are known t o be c o n s e r v a t i v e . Apparent pressure diagrams, w i t h maximum pressures less than those g i v e n by Peck, were d e r i v e d f o r American and B r a z i l i a n r e s i d u a l s o i l s (see S e c t i o n 7 . 2 . 4 ) . However, t h e r e are no r e p o r t e d f i e l d measurements o f s t r u t loads t o e s t a b l i s h t h e apparent pressure diagrams f o r Hong Kong r e s i d u a l s o i l s and s a p r o l i t e s . Despite its l i m i t a t i o n s (see Section 5 . 4 ) , t h e 'beam on e l a s t i c f o u n d a t i o n 1 approach is commonly used in Hong Kong f o r t h e design o f excavation support systems. The v a l u e s o f c o e f f i c i e n t o f subgrade r e a c t i o n recommended by Terzaghi ( 1 9 5 5 ) , although known t o be c o n s e r v a t i v e (Habibagahi & Langer, 1984), is normally employed i n t h e c a l c u l a t i o n s . The d i s c r e t e element method u t i l i s i n g t h e r e l a t i o n s h i p between l a t e r a l e a r t h pressure and displacement has a l s o been used. The boundary element method was employed t o p r e d i c t t h e performance o f t h e basement f o r t h e new Hong Kong Bank headquarters b u i l d i n g ( F i t z p a t r i c k & W i l l f o r d , 1985; Humpheson et a l , 1 9 8 6 ) . However, t h e r e are no r e p o r t e d case s t u d i e s in which t h e f i n i t e element method is used f o r t h e design o f e x c a v a t i o n support systems. I n p r e d i c t i n g g r o u n d movements around e x c a v a t i o n s , the s e t t l m & o t c u r v e s aJLJBscJc.^JLaSStL} and O ' R o u r k e e t a] (1976) (which are given i n Geoguide 1) are o f t e n used. I t T s T o u b t f u l whether these c u r v e s , which were compiled from data i n North America, are d i r e c t l y a p p l i c a b l e t o Hong Kong r e s i d u a l s o i l s and s a p r o l i t e s . / A l a r g e amount o f data have been o b t a i n e d d u r i n g t h e c o n s t r u c t i o n o f t h e Mass T r a n s i t Railway (MTR) underground s t a t i o n s and o t h e r deep basement e x c a v a t i o n s . ^ Some o f t h e MTR s e t t l e m e n t d a t a have been p u b l i s h e d (js.g. Morton et a f , J . 9 8 0 _ a & b and Budge-Re4d e t a ] , 1984). I t would t h e r e f o r e be u s e f u l t o t r y t o e s t a b l i s h s e t t l e m e n t s curves from these d a t a , s i m i l a r t o those d e r i v e d by Peck, f o r Hong Kong r e s i d u a l s o i l s and s a p r o l i t e s . An i m p o r t a n t parameter and t h e d e f o r m a t i o n o f t h e r e s t , Ko (see S e c t i o n 5 . 6 ) . i n s i t u horizontal stress in i n s i t u measurement o f Ko is
t h a t a f f e c t s t h e magnitude o f t h e s t r u t l o a d s , s o i l , is t h e c o e f f i c i e n t o f e a r t h pressure at Although t h e r e is evidence t o suggest t h a t t h e Hong Kong r e s i d u a l s o i l s and s a p r o l i t e s is low, lacking.
F a i l u r e s o f sheet p i l i n g in Hong Kong were reviewed by Malone ( 1 9 8 2 ) . Case h i s t o r i e s o f excavation support f a i l u r e s i n o t h e r c o u n t r i e s are g i v e n by Sowers & Sowers (1967), Broms & S t i 11 e ( 1 9 7 6 ) , and Daniel & Olson (1982). These s t u d i e s show t h a t f a i l u r e s o f t h e excavation support system are seldom due t o inadequacies i n d e s i g n . Instead they are o f t e n caused by poor c o n s t r u c t i o n p r a c t i c e and inadequate s i t e s u p e r v i s i o n . This i n d i c a t e s t h a t c u r r e n t design methods are adequate.
77
11.
GEOTECHNICAL CONTROL OF EXCAVATIONS IN H0N6 KONG
All private development in Hong Kong is controlled by the Building Authority under the Buildings Ordinance and its subsidiary Regulations (Government of Hong Kong, 1985). The geotechnical aspects of submissions are checked by the Geotechnical Control Office. Practice Notes listing the requirements on submissions are issued by the Building Authority to Authorized Persons and Registered Structural Engineers. The following Practice Notes are relevant to excavation submissions : (a)
PNAP : 74 - Dewatering in Foundation and Basement Excavation Works.
(b)
PNAP : 78 - Requirements for a Geotechnical Assessment at General Building Plan Stage Building (Administration) Regulation 8(1)(ba),
(c)
PNAP : 83 - R e q u i r e m e n t s for Supervision of Site Formation Works Ordinance section 17.
Qualified - Building
Practice Note PNAP : 78 lists the circumstances under which a geotechnical assessment of a site for the proposed works is required to be submitted to the Building Authority at the "general building plans11 stage, i.e. the stage at which planning and fundamental approval for building is given. One of these circumstances is when the depth of the excavation exceeds 7.5 m. This requirement permits specialist geotechnical involvement at a stage when the developer's planning is still flexible, thus avoiding possible delays due to objections or constraints when detailed engineering plans are submitted. It should be noted that, irrespective of the depth of excavation, the Building Authority can, under the Buildings Ordinance, require the details of excavation and lateral support to be submitted for agreement or approval. Where dewatering is to be undertaken to facilitate excavation works, the Practice Note PNAP : 74 lists the general requirements that should be considered when preparing foundation and basement excavation submissions to the Building Authority. These include the submission of the following : (a)
the details of the dewatering proposals,
(b)
an assessment of the effects of excavation and dewatering,
(c)
the method and sequence of construction,
(d)
the location and details of instrumentation for monitoring the effects of the works on adjoining buildings, streets, land and underground services, the time interval between reading of instruments and the limiting criteria for movements and groundwater level drawdown,
(e)
the ground investigation laboratory testing results,
report,
including
78
(f)
s h o r i n g / u n d e r p i n n i n g d e t a i l s , and
(g)
piezometric and s e t t l e m e n t m o n i t o r i n g
records.
P r a c t i c e Note PNAP : 83 l i s t s t h e requirements f o r t h r e e classes of q u a l i f i e d supervision of s i t e f o r m a t i o n and r e l a t e d w o r k s . S u p e r v i s i o n by a f u l l time resident engineer ( v i z . c l a s s ( i i i ) s u p e r v i s i o n ) i s normally s p e c i f i e d by t h e B u i l d i n g A u t h o r i t y f o r w o r k s w h i c h i n v o l v e deep excavations i n reclaimed ground, and excavation and l a t e r a l s u p p o r t works i n v o l v i n g c o m p l i c a t e d w o r k i n g p r o c e d u r e s and s t a g i n g o f works* The Authorized Person/Registered S t r u c t u r a l Engineer i s responsible for ensuring t h a t the required s u p e r v i s i o n is p r o v i d e d . I t should be noted t h a t a t t h e time o f p u b l i c a t i o n o f t h i s document, l e g i s l a t i o n is being f i n a l i s e d t o i n c o r p o r a t e p r o v i s i o n s i n t h e B u i l d i n g ( A d m i n i s t r a t i o n ) Regulations which r e f e r s p e c i f i c a l l y t o t h e submission of an excavation and l a t e r a l support p l a n . When e n a c t e d , these r e g u l a t i o n s w i l l empower t h e B u i l d i n g A u t h o r i t y t o r e q u i r e t h e s u b m i s s i o n of comprehensive i n f o r m a t i o n r e l a t i n g t o t h e design o f e x c a v a t i o n s and the l a t e r a l support system. A P r a c t i c e Note i s a l s o b e i n g prepared t o g i v e s i m p l e g u i d e l i n e s on t h e r e q u i r e m e n t s f o r t h e s u b m i s s i o n o f such information. For e x c a v a t i o n s c o n s t r u c t e d f o r p u b l i c a u t h o r i t i e s , checking of submissions is also c a r r i e d out by t h e Geotechnical C o n t r o l O f f i c e where t h i s is warranted in the i n t e r e s t o f p u b l i c s a f e t y .
79
12.
CONCLUSIONS AND RECOMMENDATIONS
A review of the state of the art of designing excavation support systems and of predicting ground movements around excavations leads to the following conclusions : (a)
Failures of excavation support systems are seldom due to inadequacies in design methods, but rather are caused by poor construction practice (e.g. inadequately planned working procedure and poor structural detailing) and inadequate site supervision. This indicates that current design methods are adequate.
(b)
For the design of excavation support systems and the prediction of movements around excavations in Hong Kong, the recommendations given in Geoguide 1 (GCO, 1982), the Terzaghi & Peck (1967) diagrams for strut loads, and Peck's (1969b) settlement curves respectively, are often employed. The 'beam on elastic foundation1 approach and the boundary element method are also used.
(c)
Proper evaluation of water pressure and its effects is of utmost importance in the design of excavation support systems. The groundwater flow regime can have a significant effect on the water pressures, the earth pressures and the piping and heave potential of the ground. Proper ground investigation is necessary to obtain information on the ground and groundwater conditions for design. The location, size and condition of water-carrying services in the vicinity of the proposed excavation should also be established.
(d)
Ground movements around excavations can be caused by many f a c t o r s , some of which cannot be quantified with confidence. Therefore, it is essential that a monitoring system is implemented during construction to check design assumptions and to provide warning to allow precautionary measures to be undertaken where required.
(e)
There are no reported field measurements of strut loads to establish the apparent pressure diagrams for Hong Kong residual soils and saprolites, which may be lower than those of Terzaghi & Peck (1967).
(f)
Conservative values of coefficient of horizontal subgrade reaction recommended by Terzaghi (1955) are usually employed in the 'beam on elastic foundation1 approach. Back analysis of monitoring data from excavations should be carried out to obtain more realistic values of ground stiffness for Hong Kong residual soils and saprolites.
80
(g)
Although t h e r e i s some evidence t o suggest t h a t t h e i n s i t u h o r i z o n t a l s t r e s s i n r e s i d u a l s o i l s and s a p r o l i t e s is low, measurements o f K o a r e l a c k i n g . S u i t a b l e methods f o r measuring Ko i n t h e f i e l d should be i n v e s t i g a t e d .
(h)
A l a r g e amount o f s e t t l e m e n t d a t a has been o b t a i n e d d u r i n g t h e c o n s t r u c t i o n o f t h e Mass T r a n s i t R a i l w a y underground s t a t i o n s and o t h e r deep basement e x c a v a t i o n s . I t would be u s e f u l t o t r y t o e s t a b l i s h s e t t l e m e n t curves f r o m these d a t a , s i m i l a r t o those d e r i v e d by Peck, f o r Hong Kong r e s i d u a l s o i l s and s a p r o l i t e s *
81
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99
TABLES
101
LIST OF TABLES Table No.
Page No.
1
Advantages and Disadvantages of Various Temporary and Permanent Support Systems (after Institution of Structural Engineeers, 1975) (two sheets)
103
2
Factors Involved in the Choice of a Support System for a Deep Excavation (NAVFAC, 1982b)
105
3
Design Considerations for Strutted and Tied-back Walls (after NAVFAC, 1982b)
106
4
Recommended Factors of Safety for Overall Stability of Cantilevered and Propped Halls (Padfield & Mair, 1984)
107
5
Methods for Calculating Anchor Prestress Loads
108
105
Table 2 - Factors Involved in the Choice of a Support System for a Deep Excavation (NAVFAC, 1982b)
Requirements
'
Tie-backs, raking struts or cantilevered walls
-
2J Low initial cost
Soldier pile or sheet pile walls; combined soil batter with wall
-
3/ Use as part of « ! permanent / \ structure
•biaphragm or cylinder \ pile walls
Diaphragm walls most common as permanent walls
h. Deep, soft clay / subsurface conditions
\Strutted or rakersupported diaphragm or cylinder pile walls
Tie-back capacity not adequate in soft clays
6. 'Deep. oyercon - i solidated clays \
' 7. Avoid dewatering
/
Comments
1. Open excavation area
5. Dense, gravelly sand or clay subsoils
y
Lends Itself to Use of
Soldier pile, diaphragm or cylinder pile walls
Sheet piles may lose interlock on hard driving
Struts, long tie-backs or combination tie-backs and struts
High insitu lateral stresses are relieved in overconsolidated soils. Lateral movements may be large and extend deep into soil
t Diaphragm walls, possibly I sheet pile walls in soft subsoils
Soldier pile walls are pervious
8. Minimize movements
High preloads on stiff strutted or tied-back walls
Analyze for stability of bottom of excavation
9. Wide excavation (greater than 20m wide)
Tie-backs or raking struts
Tie-backs preferred except in very soft clay s u b soils
Cross-lot struts
Struts more economical but tie-backs still may be preferred to keep excavation open
10. Narrow excavation (less than 20m wide)
107
Table 4 - Recommended Factors of Safety for Overall Stability of Cantilevered and Propped Walls (Padfield & Mair, 1984)
Design Approach A Design Approach B
Method
Recommended range Recommended minimum for moderately values for worst conservative parameters credible parameters (c\0* (c' = 0 , 0')
Comments
Temporary Permanent Temporary Permanent works works works works Effective stress
1.2)
(a) Factor on Embedment Depth, Fd
F
1.2 This method is empirical. It should always be checked against one of the other methods.
mended *"i A 2.0
Effective 1.2 to 1.5 stress 0'>3O° 1.5 0 = 20 to 30° 1.2 to 1.5 0 ' < 20° •Total stress
P
1.2 to 1.6 (usually 1.5) Not
*Total stress (b) Factor on Moments Based on Gross Pressure,
1.1 to 1.2 (usually
Effective Ml Iui stress Factor on Moments Based on Net Available Passive Resistance, • T o t a l stress Effective (e) stress Factor on Shear Strength on Both Active and Passive Sides. Fs •Total stress
1.5 to 2.0
1.0
1.2 to 1.5
2.0 1.5 to 2.0
1.0 1.0
1.5 1.2 to 1.5
1.5
1.0
1.2
1.5 to 2.0 (usually 2.0)
1.0
1.5
1.1 to 1.2 1.2 to 1.5 (usually 1.2 (usually 1.5 except for except for 0' > 30° 0 ' > 3 O ° when lower when lower value may value may be used) be used)
1.0
1.2
These recommended Fp values vary with 0* to be generally consistent with usual values of Fs and Fr .
2.0 1.3 to 1.5 (usually 1.5)
Not yet tested for cantilevers. A relatively new method with which little design experience has been obtained.
2.0 1.2 The mobilised angle of wall friction, 6m, and wall adhesion, Cwm, should also be reduced.
1.5
Legend : Totai stress factors are speculative ,and they should be treated with caution.
108
Table 5 - Methods for Calculating Anchor Prestress Loads
Method
Reference
Clough,Weber & Lamont (1972)
Grant (1985)
Hanna & Matailana (1970)
Terzaghi & Peck rules (0.4 FH)
Terzaghi & Peck rules ( 0 . 2 5 - 0 . 3 FH)
Pressures halfway between active and at - rest
Larson, Willette, Hail & Gnaedinger (1972)
Pressures between active and a t - r e s t
Liu & Dugan (1972)
15 x wall height (in psf)
Mansur & Alizadeh (1970)
McRostie. Burn & Mitchell (1972)
At - rest pressures
Terzaghi & Peck rules (0.4 VH)
Sands: NAVFAC (1982 b)
Oosterbaan & Gifford (1972)
Weber (1982)
0.4-0.5K 0 VH (rectangular distribution) Stiff to very stiff clays : 0.15-0.3FH (rectangular distribution) Soft to medium clays: 0.5-0.6FH (triangular distribution)
Active pressures
Terzaghi & Peck rules (30 x wall height (in p s f ) )
109
FIGURES
110
Ill
LIST OF FIGURES Figure No.
Page No-
1
Types of Temporary and Permanent Support Systems (after Institution of Structural Engineers, 1975) (seven sheets)
117
2
Factors Causing Collapse of Sheet Pile Walls (after Malone, 1982)
124
3
Methods for Transferring Earth Pressure Lagging to Soldier Piles (Peck, 1969b)
4
Types of Caisson Mall Drainage System (Malone, 1982)
126
5
Strutted Mall Support Systems (Clough, 1975)
127
6
Typical Connection Details for Horizontal Struts (Goldberg et al, 1976)
128
7
Typical Connection Details for Inclined Strut and Horizontal Waling (Goldberg et al, 1976)
129
8
Prestressing Details for Struts (Goldberg et al, 1976)
130
9
Prestressing of Pipe Strut at a Corner Using Brackets as Reaction (Goldberg et al, 1976)
131
10
Tie-back Support Systems (Clough, 1975)
132
11
Nonlinear Relationship for Soil behind Flexible Retaining Structure (Haliburton, 1968)
133
12
Nonlinear Relationship for Soil below Excavation Level of Flexible Retaining Structure (Haliburton, 1968)
134
13
Maximum Percent Deviation of Actual Stress Distribution from Uniform Pressure Distribution of Plane Strain Assumption (Tsui & Clough, 1974)
135
14
Correlation between Young's Modulus and SPT fN' Value (Chan & Davies, 1984)
136
15
Approximate Stress Paths Followed by Normally Consolidated and Overconsolidated Soil Elements Behind, and in Front of, a Retaining Wall (Padfield & Mair, 1984)
137
16
Fixed-earth Support Conditions for a Cantilevered Wall (Padfield & Mair, 1984)
138
17
Methods of Basal Heave Analysis in Cohesive Soils (Clough et al, 1979)
139
from
125
112
Pa
Figure No.
9e No-
18
Ratio of Factor of Safety against Basal Heave for Anisotropic Soil to that for Isotropic Soil for Wide Excavation (B > 15 m) (Clough & Hansen, 1981)
140
19
Basal Heave Analysis in Conesionless Soils {NAVFAC, 1982b)
141
20
Penetration of Cut-off Wall to Prevent Hydraulic Failure in Homogeneous Sand (NAVFAC, 1982a)
142
21
Penetration of Cut-off Wall to Prevent Hydraulic Failure in Stratified Soil (NAVFAC, 1982a)
143
22
Calculation of Maximum Bending Moment in a Cantilevered Wall (Padfield & Mair, 1984)
144
23
Pore Water Pressure Distribution across a Retaining Wall under Steady-state Seepage Condition (Padfield & Mair, 1984)
145
24
Groundwater Flow Patterns and Resultant Water Pressures behind Excavations (Kaiser & Hewitt, 1982)
146
25
Free-earth Support Conditions for a Propped Wall (Padfield & Mair, 1984)
147
26
Calculation of Maximum Bending Moment in a Propped Wall (Padfield & Mair, 1984)
148
27
Effect of Wall Stiffness on Bending Moment and Prop Force of a Propped Retaining Wall (Potts & Fourie, 1985)
149
28
Deflection of a Sheet Pile Wall and Redistribution of Active Earth Pressure (Bjerrum et al, 1972)
150
29
Calculation of Embedment Depth of Sheet Pile Wall below Excavation Level (NAVFAC, 1982b)
151
30
Apparent Pressure Diagrams for Computing Strut Loads in Strutted Excavations (Peck, 1969b)
152
31
Calculation of S t r u t Loads in Excavation in Soft Clay (Henkel, 1971)
153
32
Apparent Earth Pressure Diagrams for Measured S t r u t Loads in Excavations in L a t e r i t i c and Weathered Sedimentary Soils (Massad e t a l , 1985)
154
33
Calculation o f Embedment Depth of Sheet P i l e Wall in Relatively Uniform Competent S o i l Conditions (Goldberg et a l , 1976)
155
113
Figure No.
Page No.
34
Ground and Support Reaction Concept for a Strutted Excavation (Eisenstein & Negro, 1985)
156
35
Comparison between Tied-back and Strutted Walls (Broms & Stille, 1976)
157
36
Failure Mechanisms of Anchored Sheet Pile Walls (Broms & Stille, 1976)
158
37
Stability Number versus Depth Parameter for Three Conditions of Equilibrium : Overturning, Sinking and Soil Heaving (Browzin, 1981)
159
38
Design Chart for Calculating Embedment Depth of a Propped Wall in Cohesive Frictional Soils (Browzin, 1983)
160
39
Values of k for Correcting Depth Parameter p for Anchor Location /M (Browzin, 1983)
40
Design Chart for Calculating Embedment Depth of a Propped Wall in Cohesionless Soils (Browzin, 1983)
41
Idealised Force System on an Anchored Wall (Hanna, 1968)
42
Prestress Design Pressures Reported for Tied-back Wall (Clough, 1975)
43
Stress Paths for Soil Elements near Excavation (Lambe, 1970)
I65
44
Effects of Wall Stiffness and Support Spacing on Lateral Wall Movements (after Goldberg et al, 1976)
I66
45
Relationship between Maximum Settlement and Stability Number for Different Batters (after Clough & Denby, 1977)
l67
46
Observed Settlements behind Excavations (after Peck, 1969b)
47
Summary of Settlements Adjacent to Strutted Excavations in Washington, D.C. (O'Rourke et al, 1976)
169
48
Summary of Settlements Adjacent to Strutted Excavations in Chicago (O'Rourke et al, 1976)
170
49
Horizontal Strains Associated with Various Stages of Strutted Excavation Construction {O'Rourke, 1981)
171
1
•ICO lo °
114
Figure No.
No
f *
50
Ratio of Horizontal to Vertical Ground Movement as Function of Coefficient of Deformation of Fixed-end Walls (O'Rourke, 1981)
172
51
Relationship between Factor of Safety against Basal Heave and Nondimensional Maximum Lateral Wall Movement for Case History Data (Clough et al, 1979)
173
52
Relationship between Maximum Ground Settlements and Maximum Lateral Wall Movements for Case History Data (Mana & Clough, 1981)
173
53
Observed Movements of Tied-back Wall Systems (Clough, 1975)
174
54
Relationship between Prestress Pressure and Tied-back Wall Movement for Sands (Clough, 1975)
175
55
Relationship between Prestress Pressure and Tied-back Wall Movement for Stiff Clay (Clough, 1975)
176
56
Maximum Building Settlements due to Slurry Trench Excavation as a Function of Foundation Depth (Cowland & Thorley, 1984)
177
57
Relationship between Total Building Settlements and Depth Factor at Hong Kong Mass Transit Railway Stations (after Morton et al, 1980a)
178
58
Maximum Building Settlement due to Diaphragm Wall Installation (Budge-Reid et al, 1984)
179
59
Average Rate of Building Settlement for Each Metre Drop in Piezometric Level for Various Foundation Systems and Subsoil Conditions (after Budge-Reid et al, 1984)
180
60
Settlement of Buildings during Excavation of Station Box Resulting from Lateral Wall Movement (Budge-Reid et al, 1984)
181
61
Semi-empirical Method to Estimate Settlement (Bauer, 1984)
182
62
Analytically Defined Relationship between Nondimensionalized Maximum Lateral Wall Movement and Factor of Safety against Basal Heave (Mana & Clough, 1981)
183
63
Analytically Defined Relationship between Maximum Lateral Wall Movement/Ground Settlement and Factor of Safety against Basal Heave (Mana & Clough, 1981)
184
115
Figure No.
Page No.
64
Effect of Wall Stiffness on Maximum Lateral Wall Movement/Ground Settlement (Nana & Clough, 1981)
184
65
Effect of Strut Stiffness on Maximum Lateral Wall Movement/Ground Settlement (Mana & Clough, 1981)
185
66
Effect of Depth to Firm Layer on Maximum Lateral Wall Movement/Ground Settlement (Mana & Clough, 1981)
185
67
Effect of Excavation Width on Maximum Lateral Wall Movement/Ground Settlement (Mana & Clough, 1981)
186
68
Effect of Strut Preload on Maximum Lateral Wall Movement/Ground Settlement (Mana & Clough, 1981)
186
69
Effect of Modulus Multiplier on Maximum Lateral Wall Movement (Mana & Clough, 1981)
187
70
Envelopes to Normalized Ground Settlement Profiles (Mana & Clough, 1981)
187
116
117
118
119
Permanent structural frame ^sssv W usss W um^umssuggg
Permanent retaining wall which provides lateral support during excavation, e.g. diaphragm wall »-
Temporary struts enable passive soil buttress to be removed before structural frame is completed ( e ) Permanent Wall Constructed Prior to Excavation (Diaphragm Wall or Contiguous Bored Pile W a l l ) and Braced from Central Permanent Construction
-Diaphragm wall, bored piling or sheet piling -Ground anchor installed during excavation
(f) Ground Anchors
Note
:
Advantages and disadvantages of these systems ore given in Table 1.
Figure 1 - Types of Temporary and Permanent Support Systems (after Institution of Structural Engineers, 1975) (Sheet 3 of 7]
128
129
130
131
132
133
134
135
136
300 Cont. Plaza building residual soil .(Martin, 1977)
E 2
Derived from pumping tests
He Normally loaded sands
o
Derived from plate and pile tests, Hong Kong 'PhilcoxJ-962) Bank of China building - Virgin compression Tomlinson, 1953)
50
c 7/
oo
= ION
o LU D O c I
•Residual soil [Martin, 1977) 10
Bank of China , plate test (Tomlinson, 1953)
50
100
150
200
250
SPT Blowcount (N blows/300 mm)
Figure 14 - Correlation between Young's Modulus and SPT 'W Value (Chan & Davies, 1984)
13?
138
139
140
Anisotropic Strength Ratio. K s = ^"ao
Figure 18 - Ratio of Factors of Safety against Basal Heave for Anisotropic Soil to that for Isotropic Soil for Wide Excavation (B > 15 m) (after Clough & Hansen, 1981)
141
142
143
(a) Coarse Sand Underlying Fine Sand Presence of coarse layer makes flow in the fine material •IT 1 more nearly vertical and generally increases seepage gradients in the fine material compared to the homogeneous cross1 Fine sections of Figure 20. X sand . If top of coarse layer is below toe of cut-off wall at a Y • depth greater than width of excavation, safety factors of .w. . X Fine sand • "O Figure 20(a) for infinite depth apply. If top of coarse layer is below toe of cut-off wall at a depth less than width of excavation, then uplift pressures are X* • Coarse sand *.£ greater than for the homogeneous cross-sections. If permeability of coarse layer is more than ten times that of fine layer, Impervious failure head (Hw) = thickness of fine layer (H2).
Cut-off w 7
X Coarse sand • y . Coarse . -a • sand • * Fine sand"'. X*
(b) Fine Sand Underlying Coarse Sand Presence of fine Iqyer constricts flow beneath cut off wall and generally decreases seepage gradients in the coarse layer If top of fine layer lies below toe of cut-off wall, safety factors are intermediate between those derived from Figure 20 for the case of an impermeable boundary at (i) the top of fine layer, and (ii) the bottom of the fine layer assuming coarse sand above the impermeable boundary throughout. If top of fine layer lies above toe of cut-off wall, safety factors of Figure 20 are somewhat conservative for penetration required.
Impervious
(c) Very Fine Layer in Homogeneous Sand If top of very fine layer is below toe of cut-off wall at a depth greater tljidn width of excavation, safety factors of Figure 20 assuming impermeable boundary at top of fine layer apply. If top of very fine layer is below toe of cut-off wall at a depth Homogeneous less than width of excavation, pressure relief is required * sand . * . *, • so that unbalanced head below fine layer does not exceed Very fine layerTclay soil) height of soil above base of layer. / X ss Im s pervious
To avoid bottom heave when toe of cut-off wall is in or through the very fine layer, (FSH3 + FCH5) should be greater 1 than FWH4. V$ = saturated unit weight of the sand Vc = saturated unit weight of the clay Vw = unit weight of water If fine layer lies above subgrade of excavation, final condition is safer than homogeneous case, but dangerous condition Homogeneous sand excavation above fine layer and pressure Very fine layer (day soil) may arise during as in the precedinc
Fiqure 21 - Penetration of Cut-off Wall to Prevent Hydraulic Failure in Stratified Soil, (after NAVFAC, 1982a)
144
146
147
148
149
M = Maximum observed bending moment MLE= Maximum bending moment from a limit equilibrium calculation P - Wall flexibility number
2£0 200 160
;
«iw
.
. see also Section 8.1.2 equation [6))
120 80 40 0 -1-5
-0-5
0-5
1-5
log p ( a ) Variation in Bending Moment with Wall Stiffness ( F r * 2 ) P = Maximum observed prop force PLE= Maximum prop force from a limit equilibrium calculation
350 300 250 -
200 ^ 150 100 50
-1-5
-0-5
0-5
1-5
log p (b) Variation in Prop Force with Wall Stiffness
(Fr=2)
Notes : (1) For definition of F r . refer to Table L. (2) P is in units of - j ^ . Figure 27 - Effect o f Wall Stiffness on Bending Moment and Prop Force of a Propped Retaining Wall (Potts & Fourie, 1985)
150
151
152
Struts
v/////////.
0.65KQ?H
Ka = t a n 2 1 4 5 " ! )
(a) Sands
1.0KVH
K = 1-m
VH m= 1-0 except where the excavation is underlain by deep soft normally consolidated clay, then m = CH
(b) Soft to Medium Clays
(c) Stiff-fissured Clays
Figure 30 - Apparent Pressure Diagrams for Computing Strut Loads in Strutted Excavations (Peck, 1969b)
154
0-08 VH
0-05
0-10VH -i—r
I
J !
"a 0-5H ES6 {059HH j a
U j:E(0S 1) -57H
rj !!fi E 2)j (0.5S 0H !i 1-0H
(a) Apparent Earth Pressures
(b) Suggested Envelope
Legend: ES Experimental section in H Depth of excavato in { ) Dsitance of the centre of pressure from the botom of excavato Figure 32 - Apparent Earth Pressure Da i grams for Measured Strut Loads in Excavations in Lateritic and Weathered Sedm i entary Soils (Massad et al, 1985)
155
Position of lowest strut Bottom of excavation
Active
Legend Ka Pt Notes:
Active earth pressure coefficient Passive earth pressure coefficient From apparent earth pressure diagram (Figure 30) (1) Compute Re = 0-5PtL. (2) Compute depth X such that Pp = Re «• Pa. w i t h minimum factor of safety of 1.5 for passive coefficient, i.e. K'p< -r^r . (3) Check Mmax < yield moment of sheeting. (4) Drive to depth d = 1-2 X .
Figure 33 - Calculation of Embedment Depth of Sheet Pile Wall in Relatively Uniform Competent Soil Conditions (Goldberg et al, 1976)
157
CO
CD
43
c
o
o CO
o ! no CD
O)
158
I (a) Penetration Failure
(b) Toe Failure
(c) Slope Failure Soil Failure
/ v
(d) Rupture of Anchors
(e) Yielding of Sheet Piles Structural Failure
Figure 36 - Failure Mechanisms of Anchored Sheet Pile Walls (Broms & Stille, 1976)
159
1.3 1.2
c J X X =L
Ax = Horizontal component of anchor load Ap = Anchor load Nomenclature
2
3
i M
Depth Parameter . P = 1 + -r Anchored Walt Legend : Anchor pull Overturning
Sinking Heave
Fiaiire 37 - Stability Number versus Depth Parameter for y Three Conditions of Equilibrium : Overturning, Sinking and Soil Heaving (Browzin, 1981)
160
161
As. 0.1 0.2 0.3 0.4 0.5 0.6 VH A =1.2 H 0.9 0 0.8 1 0.7 1 1 0.6 1 1 1 0.5 0 0 1 1 1 2 A. = 1.ii 0.9 0 0.8 0 0 1 1 1 1 0.7 1 1 1 1 1 2 0.6 1 1 1 2 2 2 0.5 1 1 2 2 3 3 A. =1-6 0-9 0 0 0 1 1 1 0.8 1 1 1 1 1 1 0.7 1 1 2 2 2 2 0-6 2 2 2 3 3 3 0-5 2 3 3 4 4 A. =1.8 0.9 0 1 1 1 1 1 0-8 1 1 1 1 2 1 0-7 2 2 2 2 3 2 0-6 2 3 3 4 3 0.5 3 4 4 5 6 5 A. =2.0 0-9 1 1 1 1 1 1 0.8 1 1 2 2 2 1 0-7 2 2 3 3 3 2 4 3 0-6 3 3 0.5 4 5 6 6 6 4 A. = 2.U 0.9 1 1 1 1 1 0 0.8 2 2 2 2 2 1 0-7 3 3 3 3 1 0-6 4 5 5 5 5 2 0.5 6 7 8 8 7 2 As. 0.1 0.2 0-3 0.4 0.5 0.6 VH Note
0.7 0.8
1 1 2 2
0 1 1 1 2
0 1 0 2 2 3 0 1 1 2 2 0 0 0 0 0 0
0.1 0.2 0.3 0.4 0.5 0.6 X - 2.8 1 1 1 1 1 0 2 2 2 2 2 0 3 4 4 4 3 0 5 6 6 6 4 0 7 8 9 9 6 0 X - 3.2 1 1 1 1 1 1 2 2 3 2 2 4 4 4 4 3 6 6 7 8 4 9 10 10 9 6 X = 36 1 1 1 1 0 3 3 3 2 1 4 5 5 4 2 7 7 7 6 3 10 11 11 10 4 X = 40 1 1 1 1 0 3 3 3 2 1 5 5 5 4 1 7 8 8 8 2 11 12 12 10 3 X = 5.0 1 1 1 1 0 3 3 3 2 0 5 6 5 4 0 9 9 8 6 13 14 13 9 X -6.0 2 2 1 1 3 4 3 2 6 6 6 3 10 10 9 5 15 15 13 7 0-1 0-2 0.3 0.4
0.1 0.2 0.3 0.4 X= 7 2 2 1 1 4 4 3 1 6 6 6 2 10 10 9 4 16 16 14 6 A. = 8 2 2 1 0 4 4 3 1 7 7 5 2 11 11 9 3 18 17 13 4 I = 10 2 2 1 4 4 3 7 7 5 12 11 8 19 18 12 A. :=12 2 2 4 4 8 7 13 12 21 19 X - 18 2 2 5 4 8 7 14 12 23 19 X =20 2 5 9 14 24 0-1
See Figure 38 for nomenclature Figure 39 - Values of k for Correcting Depth Parameter P for Anchor Location^ (Browzin, 1983)
162
163
Ground surface Excavation Stage A
J
Anchors C
A
The Wall System
Excavation Stage
N , rN2 r
Ni [^—Side friction ( t ) 1
I
if
V, = PE1 * 2TI
Notes :
H V2 = PE, • ETl
V1+V2
ETl
(1) N1 + N2+ N3 = Area under design pressure diagram. (2) Values of P^ , P2 and P3 may change as excavation continues. (3) Test loading of anchor will temporarily modify load distribution,
Figure 41 - Idealised Force System on an Anchored Wall (Hanna, 1968)
165
Note: Initially there is no flow and ultimately there is steady-state seepage.
Stresses and Strains for Soil Elements near an Excavation
Soil Element A
Soil Element B BOBO
Initial (Static) Pore Pressure, us At Ass
BiB,
Pore Pressure upon Unloading
Decreases
Decreases
Pore Pressure during Consolidation
Decreases
increases
Strain upon Unloading
Vertical compression
Vertical extension
Strain during Consolidation Undrained Shear Strength during Consolidation
Vertical compression
Vertical extension
Increases
Decreases
Pore Pressure at Steady - state Flow, uSs
Figure 43 ~ Stress Paths for Soil Elements near Excavation (Lambe, 1970}
103
102
10
105
10*
10 Me>veme»nts \generall) > 1\>mm
I I => i
>^
+0**
^ —
^ " MtDverne nts ger»erall y < 25 mm
Flexible
Stiff (kN/m3)
Legend : h
Vertical distance between support levels or between support level and excavation base Moment of inertia of wall Modulus of elasticity of wall material
per unit length
F i g u r e 44 - E f f e c t s o f Wall S t i f f n e s s and Support Spacing on L a t e r a l Wall Movements ( a f t e r Goldberg e t a l » 1976)
168
Distance from Excavation Maximum Depth of Excavation 5 -1
c o a a u X UJ
in
CO E 3 X o 2:
Notes: (1 ) Zone I -Sand and soft to hard clay, average workmanship. (2) Zone n-(a) Very soft to soft clay. (i) Limited depth of clay below bottom of excavation (ii) Significant depth of clay below bottom of excavation but Nb<5.U. (b) Settlements affected by construction difficulties. (3) Zone II-Very soft to soft clay to a significant depth below bottom of excavation and with Nb>5-1A , yu where Nb = -7~ and Cub is as defined in Figure 31. (4) The data used to derive the three zones shown in this figure are taken from excavations supported by soldier piles or sheet piles with cross-lot struts or tie-backs. Figure 46 - Observed Settlements behind Excavations (after Peck, 1969b)
169
171
Lateral wall displacement ( mm ) 30 20 10 0
Lateral wall displacement (mm) 30 20 10 0
Stage 1
Stage'
-to
Medium sand and gravel
Stiff clay
-Horizontal soil strain contour (interval =0.5x10 3 )
Lateral displacement profile
10 £
Interbedded dense sand and stiff clay
Stage 2-
Dense sand and gravel
(b) Stage 2: Excavation to Subgrade
(a) Stage 1: Initial Excavation
20 25
Hard clay
Lateral wall displacement (mm J
30 20 10 0
Maximum displacement from wall rotation and translation
1 0
Stage 2-
s
-1.0 Stage 3-
-Maximum displacement from inward bulging
10 I 15 8" Q
20 25
(c) Stage 3: Removal of Struts
(d) Principal Components of Wall Movement
Legend : y
Soil movement vector
L.ILJ0 0
\
Vector scale ( mm ) Horizontal scale t m )
Figure 4 9 * Horizontal Strains Associated with Various Stages of Strutted Excavation Construction (O'Rourke, 1981)
172
Upper level struts Sw CD =
Sw
Sw
+
S w
20 Lateral profile
displacement at
edge
of
excavation
1.5
2% 1.0
X Q) • s i
0.5
Legend: a Model test after Milligan (19831
CC Q
0.2
0.4
0.6
0.8
1.0
Coefficient of Deformation, Cp
Case
Maximum Symbol Depth (ml
Support
Soil
Location
18
Soldier piles & lagging, 5 strut levels
Sand and stiff clay
Washington, D.C.
13
Soldier piles & lagging, 3 levels of struts and rakers
Soft to medium clay
Chicago, III.
Sheet pile, 3 raker levels
Soft to medium clay
Chicago, III.
Slurry wall, 2 levels of tiebacks and rakers
Soft to medium clay
Chicago. III.
Soldier piles & lagging, 2 raker levels
Soft to medium clay
Chicago, III.
Sheet pile, 2 raker levels
Soft to medium clay
Chicago, III.
Sheet pile, 3 levels of struts and rakers
Soft to medium clay
San Francisco California
13
U
Figure 50 - Ratio of Horizontal to Vertical Ground Movement as Function of Coefficient of Deformation of Fixed-end Walls (0'Rourke, 1981)
177
178
186
187
3.0 2.0 1.5 o o 00 II o g JZ
1.0 0.8
*
-F0 S ^ 1.0 N
0.6
FOS• >1.5I
0.4
0.2 100
200 300 400 600 800 1200 2000 Modulus Multiplier, M(=-J-)
Figure 69 - Effect of Modulus Multiplier on Maximum Lateral Wall Movement (Mana & Clough, 1981)
Distance from Excavation. x Max. Depth of Excavation, H 0.5
1.0
1.5
2.0
2.5
3.0
FOS > 1.5 ^- X //Jk
Figure 70 - Envelopes to Normalized Ground Settlement Profiles (Mana & Clough, 1981)
Av
188
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Geology of Sha (1986), 85 p.
Geological Memoir No. 2
Geology of Hong Kong Island and Kowloon, by P.J. Strange & R. Shaw (1986), 134 p.
Tin,
by
R.
Addison
191 Geolog ical Memoir No. 3
Geology o f t h e Western New Territories, by R.L. Langford, K.W. Lai, R.S. Arthurton & R. Shaw (1989), 140 p.
Geological Memoir No. 4
Geology of Sai Kung and Clear Water Bay by P.J. Strange, R. Shaw & R. Addison (1990), in press.
Geolog i c a l Memoir No. 5
Geology of the North Eastern New Territories, in preparation.
Geolog ical Memoir No. 6
Geology of Lantau, in preparation.
Map HGM 20, Sheet 2
San Tin : Solid and Superficial Geology (1:20 000 map) (1989), 1 map.
Map HGM 20, Sheet 3
Sheung Shui : Solid and Superficial G e o l o g y ( 1 20 000 map), in preparation.
Map HGM 20, Sheet 4
Kat 0 Chau : Solid and Superficial Geology ( 1 :20 000 map), in preparation.
Map HGM 20, Sheet 5
HK$97
(US$17)
HK$20
(US$3.50)
Tsing Shan (Castle Peak) : Solid and Superficial Geology (1:20 000 map) (1988), 1 map.
HK$20
(US$3.50)
Map HGM 20, Sheet 6
Yuen Long : Solid and Superficial Geology (1:20 000 map) (1988), 1 map.
HK$20
(US$3.50)
Map HGM 20, Sheet 7
Sha Tin : Solid and Superficial Geology (1:20 000 map) (1986), 1 map.
HK$20
(US$3.50)
Map HGM 20, Sheet 8
Sai Kung Peninsula : JSolid and Superficial Geology (1:20 000 map) (1989), 1 map.
HK$20
(US$3.50)
Map HGM 20, Sheet 9
Tung Chung : solid and Superficial G e o l o g y ( 1 : 2 0 000 map), in preparation.
Map HGM 20, Sheet 10
S i l v e r Mine Bay : Sol id and Superficial Geology (1:2.0 000 map) , in preparation.
Map HGM 20, Sheet 11
Hong Kong and Kowloon : Solid and Superficial Geology (1:20 000 map) (1986), 1 map.
HK$20
(US$3.50)
Map HGM 20, Sheet 12
C l e a r Water Bay : S o l i d and Superficial Geology (1:20 000 map) (1989), 1 map.
HK$20
(US$3.50)
192 Map HGM 20, Sheet 13
Shek Pik : Solid and Superficial Geology ( 1 : 2 0 000 map), i n preparation.
Map HGM 20, Sheet 14
Cheung Chau : Solid and Superficial Geology ( 1 : 2 0 000 map), in preparation.
Map HGM 20, Sheet 15
Hong Kong South and Lamma Island : Solid and Superficial Geology (1:20 000 map) (1987), 1 map.
HK$20
(US$3.50)
Map HGM 2 0 , Sheet 16
Waglan Island : Solid and Superficial Geology (1:20 000 map) (1989), 1 map.
HK$20
(US$3.50)
Map HGP 5 , Sheet 6-NW-B
Yuen Long : Marble in the Designated Area Northwest New Territories (1:5 000 map) (1988), 1 map.
Free
Map HGP 5A, Sheet 2-SW-C
Deep Bay : Superficial Geology (1:5 000 map) (1989), 1 map.
HK$10
Map HGP 5B, Sheet 2-SW-C
Deep Bay : Solid Geology (1:5 000 map) (1989), 1 map.
HK$10
Map HGP 5A, Sheet 2-SW-D
Shan Pui : Superficial Geology (1:5 000 map) (1989), 1 map.
HK$10
Map HGP 5B, Sheet 2-SW-D
Shan Pui : Solid Geology (1:5 000 map) (1989), 1 map.
HK$10
US$ prices shown above are for overseas orders and are inclusive of surface postage to anywhere in the world. ORDERING INFORMATION IS GIVEN ON THE REVERSE SIDE OF THE TITLE PAGE.
